· Project File No. WX17918 Amec Foster Wheeler Preliminary Design Geotechnical Report Environment...

Submitted to:

Tetra Tech WEI Inc.Winnipeg, Manitoba

Submitted by:

Amec Foster Wheeler Environment & Infrastructure

A Division of Amec Foster Wheeler Americas Limited

440 Dovercourt Drive

Winnipeg, Manitoba

R3Y 1N4

Canada

File number: WX17918

30 September 2016

Project File No. WX17918 Amec Foster Wheeler

Preliminary Design Geotechnical Report Environment & Infrastructure

Saskatchewan Avenue over Sturgeon Creek Culvert Replacement

Winnipeg, Manitoba

30 September 2016

17918 Geotechnical Report_Preliminary Design_final.docx Page i

TABLE OF CONTENTSPAGE

4.1 Stratigraphy .........................................................................................................6

4.2 Auger Refusal ......................................................................................................9

4.3 Groundwater and Sloughing Conditions...............................................................9

6.1 Spatial Variability in Soil and Groundwater Conditions.......................................11

6.2 Foundation Strength and Deformability ..............................................................12

6.3 Artesian Groundwater ........................................................................................12

8.1 General Evaluation ............................................................................................14

8.2 Driven Steel Piles...............................................................................................14

8.2.1 General Discussion...............................................................................14

8.2.2 Geotechnical Bearing Resistance .........................................................15

8.2.2.1 Strength Limit State ..............................................................15

8.2.2.2 Service Limit State................................................................17

8.2.3 Tensile (Uplift) Resistance ....................................................................18



8.2.4 Lateral Pile Resistance .........................................................................19



8.2.4.3 Lateral Pile Analysis Results for Select Piles Sizes...............21

8.2.4.4 Inclined (Battered) Piles........................................................23

8.2.5 Minimum Embedment Depth ................................................................23

8.3 Pile Group Effects ..............................................................................................24

8.4 Artesian Groundwater Impact on Semi-Integral Abutments................................24

10.1.1 Earth Pressure Coefficients ..................................................................26




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10.1.2 Calculation of Earth Pressure Distribution and Surcharge Loads..........27

10.1.2.1 Moderate to Well Compacted Backfill Case ..........................27

10.1.2.2 Surcharge Loads ..................................................................27

10.1.3 Load Factors.........................................................................................28

11.1 Frost Penetration Depth.....................................................................................28

11.2 Pile Foundations ................................................................................................28

11.2.1 Frost Heave..........................................................................................28

11.2.2 Adfreeze Stresses ................................................................................29

12.1 Design Philosophy .............................................................................................29

12.1.1 Design Criteria......................................................................................30

12.1.2 Methodology and Model Geometry .......................................................30

12.1.3 Soil Stratigraphy and Soil parameters...................................................30

12.1.4 Piezometric Conditions and Creek Levels.............................................31

12.2 Slope Stability Results .......................................................................................32

12.2.1 Headslope Stability ...............................................................................32

12.2.2 Grouted Rip Rap Slopes.......................................................................33

12.2.3 Cofferdam and Creek Dewatering (Short Term Stability) ......................34

12.3 Slope Stability Conclusions and Recommendations for Detailed Design............35

14.1 Pavement Design Methodology .........................................................................36

14.2 Design Vehicle and Traffic .................................................................................36

14.3 Subgrade Resilient Modulus ..............................................................................37

14.4 Granular Base Course and Subbase Course Materials......................................37

14.5 Subgrade Preparation........................................................................................37

14.6 Asphalt Concrete Pavement (ACP) Alternative ..................................................39




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LIST OF TABLES

Table 4-1: Summary of Existing Pavement and Gravel Thickness ..............................................7Table 4-2: Summary of Atterberg Limit and Particle Size Analysis Results .................................8Table 4-3: Summary of Unconfined Compressive Strength Tests ...............................................8Table 4-4: Observed Slough and Groundwater Conditions During Drilling ..................................9Table 5-1: Recommended Geotechnical Resistance Factors for Driven Steel Piles ..................11Table 7-1: Factor of Safety Against Basal Heave versus Sturgeon Creek Water Level .............13Table 8-1: Recommended Parameters for Tensile Resistance of Driven Piles..........................19Table 8-2: LPile Input Parameters for Lateral Pile Analysis .......................................................20Table 8-3: Top of Pile and Grade Elevation LPile Configurations ..............................................22Table 10-1: Earth Pressure Coefficients and Soil Unit Weights .................................................26Table 12-1: Summary of Slope Stability Material Parameters....................................................31Table 12-2: Summary of Slope Stability Results for Overburden Phreatic Surface

Elevation 233.8 m, and Artesian Total Head of 234.4 m in the underlyingTill ......................................................................................................................32

Table 14-1: Summary of Granular Pavement Structure Materials..............................................37Table 14-2: ACP Pavement Alternative for 90% Reliability........................................................40

LIST OF FIGURES

Figure 1: Test Hole Location PlanFigure 2: Summary of Slotted Standpipe and Vibrating Wire Piezometer MeasurementsFigure 3: Lateral Pile Analysis Results for 610x12.7 mm Pier Pipe PilesFigure 4: Lateral Pile Analysis Results for HP360x132 mm Integral Abutment H-PilesFigure 5: Lateral Earth Pressure Induced by CompactionFigure 6: Lateral Pressures Due to Surcharge Point and Line LoadsFigure 7: Pavement Serviceability Loss versus Time for Swelling and Frost

LIST OF APPENDICES

Appendix A Preliminary Drawings for Bridge Option 1 and 2 Provided by Tetra Tech

Appendix B Test Holes Logs & Explanation of Terms and Symbols

Appendix C Pavement Core Photographs

Appendix D TH01 Photographic Core Log

Appendix E Additional Lateral Pile Analysis Results

Appendix F Slope Stability Analysis Results




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30 September 2016

17918 Geotechnical Report_Preliminary Design_final.docx Page 4

As authorized by Mr. Vaibhav Banthia, P.Eng for Tetra Tech WEI Inc. (Tetra Tech), Amec

Foster Wheeler Environment and Infrastructure, a Division of Amec Foster Wheeler Americas

Limited (Amec Foster Wheeler) conducted a geotechnical investigation for the proposed

Saskatchewan Avenue at Sturgeon Creek Culvert Replacement. The purpose of the

geotechnical investigation was to establish the general soil and groundwater conditions at the

Site, and on that basis, provide geotechnical recommendations for Tetra Tech to use in

conceptual development and proposal of a minimum of two crossing alternatives to the City

of Winnipeg, followed by eventual detailed design of a final selected alternative. Structural

pavement cross-section alternatives for Saskatchewan Avenue were also to be provided.

The following geotechnical report pertains to preliminary design of selected crossing

alternative(s).

The proposed project consists of undertaking engineering, design, and contract administration

services for replacement of the existing two-cell concrete box culvert Saskatchewan Avenuecrossing of Sturgeon Creek. Replacement of the existing crossing is being undertaken for the

following reasons:

The existing culvert is considered to be in fair to poor condition with signs of concretedeterioration.

Overtopping of Saskatchewan Avenue on several occasions in the past as well as

sever scour damage on the downstream wing walls indicate the existing crossing ishydraulically deficient to handle the Sturgeon Creek flows in spring run-off and heavy

rainfall conditions.

It is anticipated flows on Sturgeon Creek will increase as future development occurs

upstream, which may result in increasingly unstable conditions for the existingSaskatchewan Avenue crossing as well as the CP Rail crossing located approximately

20 m north.

Further to replacement of the existing box culvert crossing, the project includes the following:

There is possibility of unstable creek banks, and in this regard, investigation of slopestability approximately 300 m upstream and downstream of Saskatchewan Avenue is

required to determine what measures, if any, are required to maintain stability.

There is a desire to extend the existing multi-use pathway; located south of the culverton the east bank of the creek; across Sturgeon Creek to Cavalier Avenue.

Reconstruction of Saskatchewan Avenue between Hamilton Avenue and Cavalier

Drive.

In a meeting held at Tetra Tech’s Winnipeg office on 26 June 2016, Tetra Tech presented the

City of Winnipeg with multiple bridge crossing alternatives to the City of Winnipeg. The

alternatives varied in the span configuration (i.e. single span versus 3 span) and alignment




Winnipeg, Manitoba

30 September 2016


(i.e. maintain existing alignment or realign Saskatchewan Avenue to the South). The meeting

concluded with Amec Foster Wheeler understanding the City of Winnipeg has selected the

following two options to carry further through preliminary design for final design selection.

Preliminary drawings for each of the two options are included in Appendix A, and the options

are summarized as follows:

Option 1: Realignment of Saskatchewan Avenue south of existing and construction ofa three span bridge with 5.5 horizontal to 1 vertical headslopes. Drawings indicate

integral abutments supported on vertical driven steel H-Piles and piers supported on

vertical driven pipe piles.

Option 2: Realignment of Saskatchewan Avenue south of existing and construction ofa single span bridge with 4 horizontal to 1 vertical headslopes. Drawings indicate semi-

integral abutments supported on vertical driven steel H-Piles (back row vertical, front

row inclined).

Prior to initiating drilling on any occasion, Amec Foster Wheeler notified public utility providers

(i.e. Manitoba Hydro, MTS, City of Winnipeg, etc.) of the intent to drill in order to clear public

utilities, and where required, met with said representatives on-site. Amec Foster Wheeler also

retained the services of a private utility locator to confirm clearance from privately owned

utilities near the test hole locations.

On 6 May through 11 May 2016, Amec Foster Wheeler supervised the drilling of a total of

sixteen test holes at the approximate locations illustrated in Figure 1. Test holes TH01 andTH02 comprised deep test holes advanced near the anticipated location of new bridge

abutments, south of the existing bridge abutments. TH03 through TH06 comprised test holes

advanced through the Sturgeon Creek slopes north and south of Saskatchewan Avenue to

just below the top of underlying till. The remaining ten test holes (RW-01 through RW-10)

comprised roadway (i.e. RW) holes advanced to about 2.0 m below grade to evaluated

subgrade conditions for pavement structures along Saskatchewan Avenue. All ‘TH’ test holes

were advanced by Maple Leaf Drilling Ltd. using a track mounted Mobile B54X drill rig

equipped with 125 mm diameter Solid Stem Augers & HQ Coring. All ‘RW’ test holes were

advanced by Maple Leaf Drilling Ltd. using a truck mounted CME55 drill rig equipped with

125 mm diameter Solid Stem Augers.

During drilling, Amec Foster Wheeler field personnel visually classified the observed soils

according to the Modified Unified Soil Classification System (MUSCS). Groundwater and

drilling conditions were also recorded at the time of drilling. Grab samples were collected at

selected depths from the auger cuttings, while relatively undisturbed Shelby tube samples

were also collected from TH03, TH04, and TH06 at selected depths ranging from about 3.0 m

to 6.1 m below grade. Split spoon samples of the till at depth were collected from each of

TH01 through TH06. Split spoon samples of the clay from about 1.5 m to 2.0 m below grade

were also obtained at each of the ‘RW’ holes. The in-situ relative consistency of cohesive

overburden was evaluated within all test holes using a pocket penetrometer, as well as within

the ‘RW’ test holes using standard penetration tests (SPT), where the number of blows to




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30 September 2016


drive a split spoon sampler 0.3 m into the soil was recorded. The relative consistency of the

underlying till within all ‘TH’ test holes was evaluated using SPT results. The recorded number

of blows is shown on the logs as the SPT (N) value.

All test holes were left open for approximately ten minutes after completion of drilling to

observe the short-term groundwater seepage and sloughing conditions. TH01 through TH06

were backfilled to grade using auger cuttings and bentonite. All ‘RW’ test holes were backfilled

to approximately 100 mm below existing pavement using auger cuttings, bentonite, and gravel

fill, and capped with 100 mm of asphalt cold patch.

Following completion of the field drilling program, a laboratory testing program was conducted

on selected soil samples obtained from the test holes. The laboratory testing program

completed consisted of moisture content determinations and four unconfined compressive

strength (UCS) tests completed in accordance with ASTM Standard D2166.

Detailed test hole logs summarizing the sampling, field testing, laboratory test results, and

subsurface conditions encountered at the test hole locations are presented in Appendix B.Actual depths noted on the test hole logs may vary by ± 0.3 m from those recorded due to the

method by which the soil cuttings are returned to the surface. Summaries of the terms and

symbols used on the test hole log and of the Modified Unified Soil Classification System are

also presented in Appendix B.

4.1 Stratigraphy

Consistent with the regional geology and anticipated conditions, the stratigraphy at the test

hole locations consisted of the following, in descending order from grade level:

Organic Clay (TH01 through TH06 only)

140 mm to 280 mm of Asphalt and/or Concrete Pavement underlain by gravel fill tobetween about 280 m and 380 m below pavement surface (RW holes only)

Clay (All test holes)

Glacial Till ( TH01 through TH06 only, encountered between 229.0 and 230.0 m)

Limestone Bedrock (TH01 only at elevation 217.7 m)

Generalized descriptions of each of the soil layers are provided below. Detailed descriptions

of the soil layers above are presented in the test hole logs in Appendix B.

Organic Clay

Grass surfaced organic clay was encountered at the surface of all test holes other than the

‘RW’ holes advanced through existing pavement along Saskatchewan Avenue. The organic

clay extended to between about 25 mm and 75 mm below grade (60 mm average), and was

generally described as silty with frequent rootlets, medium to high plastic, moist, firm, and dark




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30 September 2016


grey to black. It should be noted that the thickness of organic clay over the project extents

may vary from that observed at the test hole locations.

Existing Pavement

Existing pavement encountered along Saskatchewan Avenue consisted of a mix of pavement

structure including concrete only, asphalt only, or concrete pavement with asphalt overlay. A

summary of the pavement structure encountered at each test hole is provided in Table 4-1.

Photographs of each of the recovered pavement cores are presented in Appendix C.

Table 4-1: Summary of Existing Pavement and Gravel Thickness

Test Hole ID Pavement Structure Gravel Thickness (mm)

RW-01 200 mm Concrete 100

RW-0225 mm Asphalt

175 mm Concrete80

RW-03 175 mm Asphalt 200




RW-0750 mm Asphalt

230 mm Concrete30

RW-0850 mm Asphalt

230 mm Concrete90

RW-0965 mm Asphalt

215 mm Concrete100

RW-1065 mm Asphalt

165 mm Concrete50

Clay

Consistent with expected geology within and surrounding Winnipeg, overburden beneath the

existing topsoil and pavement consisted of clay. The clay was generally described as silty,

high plastic, moist, stiff to very stiff becoming firm below about 1.5 m below existing grade,

and dark grey to grey.

Atterberg Limits and Particle Size Analyses by Hydrometer Method were undertaken on a

total of seven samples obtained from various ‘TH’ and ‘RW’ test holes, the results of which

are summarized in Table 4-2. Unconfined compression strength tests were completed on

three Shelby Tube samples collected from TH03, TH04, and TH06, the results of which are

summarize in Table 4-3.




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Table 4-2: Summary of Atterberg Limit and Particle Size Analysis Results

Test HoleDepth

(m)

Elev.

(m)

Liquid

Limit

(%)

Plastic

Limit

(%)

Gravel

(%)

Sand

(%)

Silt

(%)

Clay

(%)

RW03 0.5 236.6 80 20 0.5 15.2 15.5 68.7

RW04 0.5 235.9 82 23 3.0 15.3 17.4 64.3

RW07 0.5 236.9 90 26 0.0 3.2 21.1 75.8

RW09 0.5 237.8 87 20 0.0 0.8 27.2 72.0

TH03 3.0 231.4 96 21 0.6 4.3 7.0 88.1

TH04 3.0 229.9 92 20 0.0 3.9 18.6 77.5

TH06 3.0 232.9 76 19 0.0 2.4 30.5 67.2

Table 4-3: Summary of Unconfined Compressive Strength Tests

Test HoleDepth

(m)

Elev.

(m)UCS (kPa)

Strain at 100%

of UCS

(%)

Bulk Density

(kg/m3)

Dry Density

(kg/m3)

TH03 3.0 231.4 58 3.7 1842 1259

TH04 3.0 229.9 65 6.6 1733 1134

TH06 3.0 232.9 106 2.0 1787 1212

Glacial Till

Till was encountered beneath the clay overburden at each of test hole TH01 through TH06 at

depths ranging from about 2.9 m to 6.4 m below existing grade; equivalent to between

approximate elevations 229.0 m and 230.0 m.

The till was generally described as loose to very loose within the upper 0.6 m to 1.0 m below

the surface of the till, transitioning to very dense below. Cemented till resulting in core recovery

was observed at TH01, and is shown in the photo log provided in Appendix D. The

composition of the till varied highly, both with depth within a single test hole and between the

test hole locations. In general, some till zones were dominated by a silt and sand matrix with

some gravel and trace clay. Other zones were essentially ‘clean’ with no fines, and were

dominated by ‘clean’ gravel with trace sand. Cobble and boulder sizes were also encounteredthrough the till, and even resulting in breakage of the coring bit at TH02 forcing the hole to be

abandoned.




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Bedrock

Bedrock was explored at TH01 only. A photographic log of the bedrock core obtained at TH01

is provided in Appendix D.

In summary, the bedrock was described as medium strong, moderately weathered, dark grey

to grey limestone interbedded with frequent reddish grey mudstone layers 10 mm to 75 mm

thick, and abundant subrounded to rounded clasts. Rock quality designation (RQD) values

ranged from about 63% from 0 to 1.5 m below the surface of the bedrock, to about 94% from

about 1.5 m to 3.0 m below bedrock surface.

4.2 Auger Refusal

Auger refusal occurred in TH01 at approximate elevation 228.1 m (5.3 m below grade) and in

TH02 at approximate elevation 228.3 m (5.6 m below grade) prior to switching to HQ coring.

Test holes TH03 through TH06 achieved target termination depth and were not taken to auger

refusal.

4.3 Groundwater and Sloughing Conditions

Seepage and sloughing conditions were noted during drilling, and the depth to the

accumulated water level within the test holes was measured about ten minutes after drilling

at each test hole location. Recorded observations at test hole locations TH01 through TH06

are summarized in Table 4-4. Neither seepage nor groundwater was observed in any of the

‘RW’ test holes advanced through Saskatchewan Avenue to 2.0 m below grade.

Table 4-4: Observed Slough and Groundwater Conditions During Drilling

TestHole

ID

TestHole

Depth(m)

During Drilling Upon Completion

Sloughing

Zone

Seepage

Zone

Depth to

Slough (m)

Depth to

Groundwater (m)

TH01 18.7 Below 14.6 m

Flowing artesianconditions

(approx. 4 L/min.)encountered upondrilling to 15.5 m

N/A - Cased testhole for wet coring

-0.3 (i.e. aboveexisting grade)

TH02 9.0 Below 5.5 m Below 5.5 mN/A - Cased test

hole for wet coring3.0

TH03 6.6 Below 6.4 m Below 6.4 m 5.8 4.0

TH04 4.8 Below 4.6 m Below 4.6 m 4.3 1.8

TH05 3.5 None At 3.5 m 3.5 Trace at bottom

TH06 7.2 None None 7.0 Dry

Further to seepage and groundwater observation at the time of drilling, TH01, TH02, TH04,

and TH05 were instrumented with slotted standpipe piezometers in order to allow for short

term monitoring of groundwater levels within the till (TH04 and TH05) and bedrock (TH01 and




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TH02). Vibrating Wire (VW) piezometers equipped with single channel data logger obtaining

measurements at 4 hour intervals since installation were installed within the clay overburden

at TH03 and TH06 in order to monitor porepressure within the Sturgeon Creek slopes

upstream and downstream of Saskatchewan Avenue. All recorded standpipe and VW

piezometer readings obtained through 29 June 2016 are illustrated in Figure 2.

In summary, seepage and groundwater observation during and post drilling indicate an

artesian bedrock condition with an estimated phreatic surface of up to elevation 238.0 m north

of Saskatchewan Avenue, and up to about 235.1 m south of Saskatchewan Avenue.

Per clause D5.9 of City of Winnipeg RFP 3-2016, the bridge shall be designed in accordance

with CAN/CSA-S6-06, the Canadian Highway Bridge Design Code (CHBDC), to be

structurally and functionally safe for the duration of a minimum 75 year design life.

Minimum requirements for the design of foundations and geotechnical systems are outlined

in Section 6 of the CHBDC. CHBDC employs Limit States Design principles, wherebygeotechnical resistance factors are applied to the ultimate geotechnical resistance to obtain a

factored geotechnical resistance for the specified limit state. The geotechnical resistance

factor to be applied in design is to be selected from those outlined in the CHBDC on the basis

of the degree of understanding of the site and prediction model for design. Based on the

regional geology, subsurface conditions encountered at the site, and Amec Foster Wheeler’s

experience and empirical knowledge of local foundation performance within the City of

Winnipeg, Amec Foster Wheeler recommends the geotechnical resistance factors outlined in

Table 5-1, selected from those outlined in the CHBDC. The geotechnical resistance factor to

be applied in design shall be selected in accordance with the Analysis Method / Predictive

model used in final selection of the pile driving equipment and development of driving criteria.




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Table 5-1: Recommended Geotechnical Resistance Factors for Driven Steel Piles

Limit State Analysis Method / Predictive ModelResistance

Factor

Ultimate Limit State (ULS)

Compression, ϕgu

Static Analysis without driveability analysis (WEAP),

dynamic pile measurements, or load tests.0.40

Hammer selection and driving criteria established by

driveability analysis (WEAP), without dynamic pile

measurements or load testing

0.45

Driving criteria established by driveability analysis (WEAP)

with confirmation of hammer performance with dynamic pile

(PDA) measurements and signal matching (CAPWAP)

0.50

Tension, ϕgu

Static Analysis 0.30

Dynamic Testing with Signal Matching (CAPWAP) 0.40

Lateral, ϕgu All analysis methods 0.50

Serviceability Limit State (SLS)

Settlement or

Lateral Deflection,

ϕgs

Static Analysis 0.9

Based on subsurface soil and groundwater conditions encountered at the test hole location,

the following key geotechnical considerations have been identified and will need to be

considered throughout design:

6.1 Spatial Variability in Soil and Groundwater Conditions

The soil and groundwater conditions encountered during the investigation were characterized

based on conditions observed in small diameter test holes advanced at relatively wide

spacing. Subsurface conditions at locations that were not investigated could vary from the

conditions observed in the test holes. Spatial variability of sub-surface conditions should be

expected and allowed for in the design and construction processes.

The stratigraphy and soil encountered within the test holes advanced at the site are

considered typical of geologic conditions within the region and are consistent with anticipated

subsurface conditions. However the depth to bedrock was greater than anticipated and

resulted in greater than normal/average till thickness.




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6.2 Foundation Strength and Deformability

The high plastic lacustrine clay overburden is highly deformable (both vertically and laterally)

under surcharge loads. Construction of embankment fills over the compressible native

lacustrine clay foundation soils will result in long term consolidation of the clay foundation

soils, and by extension, settlement of embankment fill. Lateral deformation of the

embankments and underlying native clay soils will also occur following loading, and may be

more pronounced below the edges of embankment areas. In this regard, selection and design

of foundations must consider the potential impact of new embankment fill on both vertical and

lateral soil deformations around the foundation.

Notwithstanding the above, based on comparisons of the preliminary top of deck elevation of

about 236.150 m to existing roadway and ditch elevations along Saskatchewan Avenue, fill

requirements are expected to be limited to less than 1 m over the project extents. This

includes any fill required to construct the headslopes and sideslopes at the proposed

abutments, for both Options 1 and 2. In this regard, long terms consolidation of foundationclays is not expected to be an issue pending confirmation of fill depths less than 1 m during

detailed design. If during detailed design fill depths greater than 1 m are identified, the impact

of these fill depths on foundation performance should be evaluated by Amec Foster Wheeler

during detailed design.

6.3 Artesian Groundwater

Observed groundwater levels within the till and bedrock and porewater pressures measured

within the clay identify phreatic surfaces above existing grade. In this regard, groundwater

conditions, at the site, in particular artesian groundwater pressure with the till and bedrock,

present a high risk for basal heave and excavation instability. The clay overburden is an

aquiclude, the removal and/or thinning of which could result in basal heave and development

of a flowing artesian condition, which would necessitate temporary depressurization of the

aquifer to complete construction and restore the aquiclude / stability. Excavations presenting

risks to the project include all temporary excavations, as well as any permanent excavations

or voids inclusive of annular voids around integral abutment piles (to allow abutment

movement due to thermal expansion and contraction). The impact of cofferdam construction

and any dewatering and/or subexcavation of the creek bed shall be considered throughout

design, particularly as it relates to constructability.

To the best achievable extent, design and construction planning should seek to keep all

temporary and permanent excavation requirements to a minimum; preferably zero. In

particular, design and construction planning should seek to eliminate subexcavation of theexisting creek bed as comparison of artesian pressures in the underlying till and bedrock

suggest that Sturgeon Creek, at this location, may be a zone of perpetual groundwater

discharge. The rate of groundwater discharge at the base of the creek is mitigated by head

loss through the clay layer between the base of the creek and underlying till.

Additional discussion and recommendations for determining the factory of safety against basal

heave is presented in Section 7.0.




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Artesian groundwater pressures with the underlying till and bedrock require that care be taken

in selection and design of the required dewatering, and excavation depths (both temporary

and permanent) in order to mitigate the risk of basal heave and loss of stability. This includes

evaluating the impact of temporary dewatering on stability of the bed of Sturgeon Creek, as

well as evaluation for the potential of developing perpetual seepage and groundwater flow

along integral abutment piles where annular voids are provided to accommodate movement

of the abutments.

According to Canadian Foundation Engineering Manual (CFEM), the porewater pressure

acting at the bottom of an aquiclude (i.e. on the underside of the highly plastic overburden)

should not exceed 70% of the total weight of soil and groundwater above this depth. In other

words, a minimum factor of safety of 1.4 against basal heave is recommended, whereby the

factor of safety is determined as the ratio of total weight above the point of pressure to the

porewater pressure acting at that point.

For evaluation purposes, the base of the aquiclude may be assumed equivalent to the surface

of the glacial till. Till surface elevations varied at the test hole locations between about 229 m

to 230 m. Based on the water levels measured at the standpipe locations (see Figure 2),

porewater pressure acting on the underside of the aquiclude may be determined assuming a

phreatic surface (i.e. total head) of 234.5 m. This equates to a pressure head of 55 kPa acting

on the underside of the aquiclude for a till surface elevation of 229.0 m, and 45 kPa for a till

surface elevation of 230.0 m.

Assuming a creek bed elevation of 231.2 m and a bulk density of 16 kN/m3 for overburden

soil comprising the creek bed, the factor of safety against basal heave as a function of till

elevation and water level in the creek is outlined in Table 7-1.

Table 7-1: Factor of Safety Against Basal Heave versus Sturgeon Creek Water Level

Water Elevation in Sturgeon CreekFactor of Safety

Till Elev. 229.0 m Till Elev. 230.0 m

231.2

(Dewatered Channel)0.64 0.35

232.67

(Water Level 22 June 2016)1.33 1.04

233.37 (Q50) 1.66 1.37

234.52 (Q1) 2.19 1.91

In summary, the results indicate that dewatering of the creek will severely jeopardize the

stability of the base of the creek against basal heave. In this regard, localized depressurization

of the aquifer beneath the creek bed would be required to ensure a stable creek bed

throughout construction. The pumping/extraction rates required to lower the groundwater

pressure head are dependent on the transmissivity of the aquifer, Completion of a




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hydrogeological study and well pumping program was beyond the scope of this report;

however, would be undertaken during detailed design in support of a depressurization system

of the underlying aquifer.

Further to stability of the creek, determination of the required overburden pressure to resist

basal heave indicates permanent excavations for the annular voids around integral abutment

piles extending below elevation 233.9 m will require some form of seal to mitigate groundwater

flow through the void space. Recommendations for design of a seal are discussed in Section

8.4.

8.1 General Evaluation

Based on preliminary design discussions with Tetra Tech, Amec Foster Wheeler understands

driven steel piles are the preferred foundation alternative. Based on the subsurface conditions

at the test hole locations and the key geotechnical considerations discussed in Section 0,

Amec Foster Wheeler supports driven steel piles as the preferred pile foundation alternativeat this site. Where employed, it is expected that driven steel piles would be driven to ‘practical

refusal’ refusal within the underlying till and/or on bedrock.

Piles at this site are expected to develop greater than 90 percent of their resistance in

combined shaft friction and end bearing within the till. The depth of penetration into the till will

be restricted by end bearing development which in turn will be highly influenced by the

presence of cobbles at boulders at the pile tip. In this regard, the variability in achieved (i.e.

as-built) pile embedment depths is expected to be greater than normal. Further, cobbles and

boulders induce severe driving conditions and high localized driving stresses within the pile

sections. As such, steel pile sections are recommended due to their ductility. Driven pre-cast

pre-stressed concrete piles are susceptible to brittle failure and/or breakage at the pile tip and

are not recommended. The use of conventionally bored friction piles is not recommended for

support of bridge structures given the limited thickness and available resistance of the clay

overburden. Furthermore, extension of bored piles into the underlying till is not recommended

due to the artesian conditions and potential development of flowing groundwater conditions

discussed in Section 6.3.

Based on the discussion above, foundation recommendations presented in this report are

limited to driven steel piles. Foundation recommendations for alternate pile types can be

provided upon request.

8.2 Driven Steel Piles

8.2.1 General Discussion

Amec Foster Wheeler understood preliminary designs propose to use driven steel H-Piles for

support of the bridge abutments, where-as pipe piles are proposed for support of piers. The

following additional comments are provided with respect to driven steel piles:




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Amec Foster Wheeler anticipates steel piles could be driven to end bearing in theunderlying silt till and/or bedrock. It is anticipated that steel piles could be driven with

relative ease through clay embankment fill and overburden; however, severe driving

conditions associated with cobbles and boulders in the till and end bearing on rock

are expected. In this regard, the toes of all driven steel piles should be equipped with

am internally fitted, hardened (i.e. cast) steel driving shoe continuously welded or

fastened to the pile in order to reinforce the toe of the pile during driving.

Both H-Pile and Pipe piles are expected to develop a plugged toe condition (i.e. endbearing applied to full cross-sectional area of pile tip) with penetration into the

underlying till.

H-Piles are anticipated to penetrate the till to grater depths than pipe piles.

Due to artesian conditions within the underlying till, the interior of pipe piles shouldbe infilled with concrete to the underside of pile cap in order to prevent loss of pile

toe support associated with upward seepage, basal heave, piping at the pile.

Where a CSP (or other material) is provided to maintain a void around integralabutment, special precautions such as a grout seal between the pile and the CSP

will be required to prevent upward seepage and groundwater flow along the pile

induced by artesian pressure in the underlying till and bedrock.

8.2.2 Geotechnical Bearing Resistance

8.2.2.1 Strength Limit State

The ultimate geotechnical resistance of a steel pile driven to ‘practical refusal’ using an

appropriately sized hammer and driving force and energy shall be limited to a maximum given

by the structural column capacity of the steel pile section, which may be taken as 0.63FyAs

for severe driving conditions, where Fy is the yield strength of the steel, and As is the cross-

sectional steel area. The geotechnical resistance factor for determination of factored bearing

resistance should be selected in accordance with the recommendations in Section 5.0 in

consideration of pile load testing undertaken at the time of construction.

Additional recommendations for design and construction of driven steel piles are as follow:

The pile capacity given by the above pertains to the geotechnical resistance of a fullyembedded with zero unsupported pile length.

Pile cross sections must be structurally designed to withstand the design loads andthe driving forces during installation. Evaluation of the structural resistance of piles

shall be undertaken by the structural engineer, and should consider laterally

unsupported pile length. Examples of unsupported pile length include pile stick-up, or

any annular void space provided around any embedded portion of the pile to preclude

lateral resistance, such as is proposed by Tetra Tech for the integral abutment

alternative.




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The effect of corrosion and deterioration from environmental conditions shall beconsidered in selection of the required pile cross-section for long term pile capacity.

The potential for corrosion and anticipated corrosion rates should be investigated

during detailed design by a corrosion expert. Amec Foster Wheeler anticipates that

corrosion of steel piles will likely be addressed by sacrificial steel thickness, although

other alternatives may be adopted.

As a minimum, steel H-piles should meet the requirements ofCAN/CSA-G40.20/G40.21, Grade 350 W, and pipe piles should have a minimum yield

strength of 310 MPa (i.e. ASTM A252 Grade 3 steel). The toes of all H-piles shall be

equipped with a hardened (i.e. cast) steel driving shoe continuously welded or

fastened to the pile in order to reinforce the toe of the pile during driving.

Wave Equation and driveability analysis (i.e. WEAP) should be completed for each ofthe selected pile sections, prior to proceeding to construction and concurrent with

selection of the pile driving equipment, to confirm the ability of the proposed hammer

and appurtenances to drive the piles to the required design capacity and embedment

depth without damage. Similarly, the WEAP shall be extended to develop termination

criteria for use in pile installation monitoring.

An appropriately selected pile driving hammer and appurtenances shall be capable ofdriving the specified pile section to the design ultimate geotechnical resistance at a

termination criterion (or pile driving acceptance criterion) defined by a penetration

resistance of not less than 5 blows per 25 mm of penetration and no more than 15

blows per 25 mm of penetration. In order to mitigate the risk of damage to both the pile

and piling equipment, continuous driving of a pile at penetration resistances exceeding

15 blows per 25 mm of penetration should be avoided.

To reduce the potential for structural damage to the pile, maximum compression andtension driving stresses developed within the pile during installation shall be limited to

≤0.9Fy for steel piles. If WEAP analysis prior to construction predicts driving stresses

exceeding 0.9Fy, then the foundation design and/or selected pile section shall be

revised until the design pile resistance is achieved within the recommended driving

stress limit.

For an appropriately sized pile driving hammer, it is anticipated that piles could be

driven through the embankment fill and clay overburden with relative ease. However,

potentially highly variable pile capacity development (i.e. driving resistance) is

anticipated over the depth of penetration into the underlying till. In particular, potential

cobbles and boulders in the till could provide for difficult driving conditions and sudden

refusal. A contingency should be carried to allow for potentially highly variable pile

lengths across the site.

Excessive penetration resistance (i.e. values greater than 15 blows per 25 mm ofpenetration) is expected to develop shortly following contact of the pile toe with the

bedrock surface. It should also be noted that potential cobbles and boulders in the till

could provide for increased penetration resistance and sudden refusal. As indicated

above, in order to mitigate the risk of damage to both the pile and piling equipment,




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continuous driving of a pile at penetration resistances exceeding 15 blows per 25 mm

of penetration should be avoided.

If damage to a pile is observed during driving, driving should cease immediately, andthe pile capacity and extent of damage assessed by a qualified geotechnical engineer

and structural engineer. This may include estimating the pile capacity and assessing

damage by dynamic testing. Any piles that have been damaged, are excessively out

of plumb, or have refused prematurely may need to be replaced pending the results of

the review.

Dynamic (PDA) testing of a minimum given by the greater of 5% of the total pile countand one pile per pier and abutment substructure unit should be undertaken to validate

the pre-construction WEAP analysis and verify that pile driven to the established

termination criteria meet design requirements. Dynamic pile measurements (PDA)

should also be used to monitor for indications of pile damage.

Prior to the pile installation, the piles should be inspected to confirm that the materialspecifications are satisfied. The piles should be free from protrusions, which could

create voids in the soil around the pile during driving.

All piles driven within five pile diameters of a previously driven pile should have thepreviously driven pile monitored for heave and, where heave is observed, the heaved

pile should be re-driven. Piles that are re-driven should be advanced to at least the

original elevation.

As driven steel pile installations do not allow for direct confirmation of soil conditionsduring construction, and the piles cannot generally be visually inspected for damage

following installation, construction monitoring will be important in quality control and

quality assurance of pile installations, and to verify that the piles are installed in

accordance with design assumptions and the driving criteria are satisfied. Construction

monitoring of pile installations should be undertaken on a full time basis, and should

include, but not be limited to, confirmation that pile materials meet or exceed

specifications, confirmation of hammer and appurtenance operating conditions, and a

detailed driving record inclusive of penetration resistance (i.e. blow counts per unit

penetration) and blow rate or energy. Completed pile driving records should be

reviewed on a regular basis during pile driving by a qualified geotechnical engineer.

8.2.2.2 Service Limit State

The settlement of a single pile depends on the applied load, strength-deformation properties

of the foundation soils, load transfer mechanism, load distribution over the pile embedment

depth, and the relative proportions of the load carried by shaft friction and end-bearing. A pile

settlement limit value was not specified by the structural agent for use in developing

geotechnical resistance limits for the serviceability limit state design criterion.

The settlement of a single pile driven to refusal in the till or on the underlying bedrock (using

an appropriately sized hammer and driving energy) is expected to be governed primarily by

elastic shortening of the pile section under applied loads. It is further anticipated that the




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majority of settlement would occur during construction under the progressive application of

sustained dead loads.

Abutments piles will be required to extend through embankment fill. Consolidation of the

embankment fill and underlying compressible clay would induce negative skin friction or drag

load on piles driven to refusal in the till or on the bedrock, and would result in additional stress

development within the pile section and increased elastic shortening of the pile. Consequently,

it is important that the potential for drag loads be considered in addition to structural loads

when estimating the settlement performance of piles extending through fill.

The estimated total pile head settlement of a steel pile driven to practical refusal in accordance

with the recommendations of this report can be estimated as a function of elastic shortening

of the pile under the applied service load plus drag load. It is recommended that the settlement

performance of the pile be approximated assuming a fully end bearing pile (i.e. zero shaft

resistance) where the drag load is assumed to be applied at the top of the pile (rather than

along the pile shaft) as follows (CFEM 3, 1992):

where: S = Total pile head settlement (m)

b/100 = Approximate toe mobilization (m)

b = Pile diameter (O.D., in m)

es = The elastic compression of the pile (m)Q = Sum of applied unfactored working load and drag load (kN)

L = Pile length (m)

A = Cross-sectional area of the pile material (m2)

E = Elastic modulus of the pile material (kPa)

Recommendations for determining the drag load are discussed in Section 9.0.

8.2.3 Tensile (Uplift) Resistance


When tensile forces are present, the ultimate tensile resistance of driven piles should be

determined using unit shaft friction values outlined in Table 8-1. For pipe piles, only the

exterior surface area of the pile in contact with the soil should be used in the calculation of thefrictional resistance. In the case of steel H piles, the surface area should include the exterior

sides of the two flanges plus twice the depth of the web. Although not commonly employed

for the installation of driven steel piles, if pre-boring is required (i.e. for ground disturbance

clearance or contractor preference), shaft friction must be neglected over the depth of the pre-

bore for H-Piles, and over the depth of the pre-bore for pipe piles if the pre-bore exceeds 95

percent the pile diameter. If the pre-bore is limited to no more than 95 percent of the pile

diameter, no reduction in shaft friction is required for pipe piles. Shaft resistance should also

AEQLbbS es

100100δ




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be neglected over the portion of the piles that may be inserted within sleeves, such as has

been proposed by Tetra-Tech for the integral abutment alternative (Option 1).

Table 8-1: Recommended Parameters for Tensile Resistance of Driven Piles

Elevation Range Anticipated Soil TypeUltimate Unit

Shaft Friction

El. 233.0 to El. 229.0 Clay Fill / Clay 30*

El. 229.0 to El. 228.0 Loose Till 30

El. 228.0 to El. 222.0 Dense Till

Linearly Increasing with depth from:

30 kPa at El. 228.0 to

90 kPa at El. 222.0

Below El. 222.0 Dense Till 90 kPa

* The upper 2.4 m of the shaft, below final grade, is to be ignored in estimating shaft friction to account for the loss of

contact between the soil and pile interface, such as may result from seasonal frost, moisture changes, etc.

The geotechnical resistance factor for determination of factored bearing resistance should be

selected in accordance with the recommendations in Section 5.0 for pile design using Static

Analysis.


The upward displacement of a pile in tension can be estimated in the same manner as

estimating the settlement of a friction pile in compression. In this regard, the upward

displacement of a driven steel pile under in tension under a maximum factored load given by

the recommendations in Section 8.2.3.2 is expected to be in the range of 0.05 to 0.2 percent

of the shaft diameter, plus elongation of the pile due to the applied tension load.

8.2.4 Lateral Pile Resistance


Piles resist laterally applied loads by deflecting until the necessary resistance is mobilized in

the adjacent soils. The lateral capacity depends upon the properties of the soil and pile

materials, pile size, fixity at the top of the piles, depth of embedment, load distribution along

the pile, and tolerable deflections. Where lateral pile resistance is required, it is recommended

that the nominal horizontal load resistance of piles be estimated using procedures that

consider soil-structure interaction, such as the method of non-linear p-y curves, whereby

horizontal resistance is estimated based on both the non-linear strength-deformation

characteristics of the soil stratum surrounding the pile and the structural properties of the pile.

Based on Reese and others (1984), the soil reaction (p) is related to the shaft deflection (y)

for various depths below the ground surface. In general, p-y curves are nonlinear and depend

on several parameters, including depth, shaft diameter and soil strength.

Based on conditions observed within the appended test hole logs, the stratigraphy and soil

parameters outlined in Table 8-2 are considered suitably representative of the average




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subsurface conditions expected to influence the lateral behaviour of driven steel piles at the

site. Revisions to the stratigraphy outlined below may be required by design and construction

requirements (i.e. at abutments where sand or gravel may be used as opposed to clay), and

as such, should be reviewed by the design engineering during design. Furthermore, Amec

Foster Wheeler should be notified of any deviations from the stratigraphy outlined in Table

8-2 for review.

Table 8-2: LPile Input Parameters for Lateral Pile Analysis

Elevation Range (m)Soil

Type

Soil

Model

Effective

Unit

Weight

(kN/m3)

Friction

Angle

(°)

Undrained

Shear

Strength

(kPa)

E50

(%)

p-y

subgrade

modulus, k

(kPa/m)

Final grade to:

underside of rip-rap

Rip-

Rap

Ignore strength contribution provided by rip-rap in the event of movement

and formation of a gap at the rip-rap / pile interface.

Lesser of: final grade

above existing or

underside of rip-rap to:

El. 234.0

Clay

Fill

Soft Clay

(Matlock)17.5 n/a 30 0.0088 Default

Lesser of: final grade

below El. 234.0 m,

underside of rip-rap

layer, or El. 234.0 m to:

El. 229.0

Clay

Fill /

Clay

Soft Clay

(Matlock)7.7 n/a 30 0.0088 Default

El. 229.0 to El. 228.0Loose

TillAPI Sand 10.2 30 n/a n/a 16,000

El. 228.0 to El. 217.0Dense

TillAPI Sand 12.2 35 n/a n/a 24,000

The recommended ‘soil model’ outlined in Table 8-2 for analysis of the lateral load-

deformation of piles have been presented for the condition of intimate contact between the

pile and surrounding soil. Lateral soil resistance shall be neglected if the soil providing

resistance is, or is likely to become soft, loose, removed due to scouring, or disturbed, or if

the contact between the soil and wall is not tight. In this regard, lateral resistance should be

neglected over the depth of seasonal frost penetration, or in the case of semi-integral

abutments, over the length of the annular void maintained around the piles to allow movement

of the abutment (i.e. in response to thermal expansion and contraction of the bridge deck).

Boundary conditions for LPile analysis should further be selected in accordance with the

functional and performance requirements, and whether or not movement of the pile head is

load controlled (i.e. prediction of pile head deflection in response to applied loads), or

deflection controlled (i.e. evaluation of shear and bending moments within the pile in response

to known movement of the pile head).

Where piles are required to provide horizontal resistance, evaluation of the nominal horizontal

load resistance of a pile or pile group at the strength limit state requires that a criterion (or

criteria) defining the strength limit state of piles under horizontal load be defined. The nominal

horizontal resistance of a pile may be defined in accordance with one of the three following

theoretical scenarios (CFEM, 2006):




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1. The capacity of the soil may be exceeded, resulting in large horizontal movements of

the piles and failure of the foundation;

2. The bending moment (M) and/or shear (V) may generate excessive bending or shear

stresses in the pile material, resulting in structural failure of the piles; or

3. The deflection of the pile head may be too large to be compatible with the

superstructure (i.e. the deflection that can be tolerated at the foundation/

superstructure interface prior to initiating the ultimate limit state in the superstructure).

Case 1 is approximately similar to loss of fixity at the pile toe, and mobilization of plastic

deformation of the soil along the full embedment length of the pile. For this determination, the

loads applied to the pile are factored, and a soil resistance factor of 0.5 is used.

For Case 2, the method of p-y curves can be used to evaluate the bending moment and shear

generated in a given pile configuration under an applied load scenario. The nominal horizontal

resistance can then be determined as the maximum load that can be applied to the pile priorto exceeding the bending moment resistance or shear resistance of the pile.

For Case 3, the lateral deflection predicted by LPile under an applied loading condition should

be compared for compatibility with the maximum horizontal movement that can be tolerated

by the superstructure at the strength limit state. Note that the level of deflection that can be

tolerated at the strength limit state may be larger than the level of deflection defining the

serviceability limit state. Where the deflection that can be tolerated by the pile exceeds the

maximum tolerable level of deflection at the strength limit state, the nominal horizontal

resistance of the pile should be limited to the horizontal load scenario corresponding to the

maximum level deflection that can be tolerated by the superstructure.


The horizontal movement of pile foundations shall be estimated using the method of non-

linear p-y curves discussed in Section 8.2.4.1. Tolerable horizontal movement of piles shall

be established on the basis of confirming compatible movements of structural components for

the loading condition being considered.

8.2.4.3 Lateral Pile Analysis Results for Select Piles Sizes

At the request of Tetra Tech, Amec Foster Wheeler undertook lateral pile analysis (using LPile

produced by ENSOFT Inc.) of the 610x12.7 mm pier pipe piles (concrete filled) and of the

HP360x132 abutment piles proposed for the three span bridge with integral abutment

alternative (Option 1). With respect to the H Piles, Amec Foster Wheeler understood analysis

was only required for flexure about the weak axis. LPile analyses were not required for Option

2 (Single Span Bridge) which employed a front row of inclined driven steel H-Piles to resist

lateral foundation loads.

Lateral loading analyses were performed for both fixed-head and free head conditions

assuming static, sustained loading conditions. With respect to the 610x12.7 mm pipe piles,

lateral displacement of the piers and piles is expected to be governed by applied loads (i.e.




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30 September 2016


load controlled boundary condition). In this regard, the purpose of LPile modelling was to

predict lateral pile movement in response to applied loads. Modelling of the free head

condition for the 610x12.7 mm pipe piles consisted of predicting the lateral deflection of the

pile for a specified combination of lateral load, vertical load, and moment applied to the pile

head. Modelling of the fixed head condition consisted of predicting the lateral deflection of the

pile for a specified combination of lateral and vertical force applied to the pile head while

maintain zero rotation of the pile head.

With respect to the HP360-132 integral abutment piles, the magnitude of lateral pile

displacement is restricted by thermal expansion and contraction of the structure (i.e.

displacement controlled boundary condition). Integral abutments seek to allow this movement

to occur with minimal soil resistance, and in this regard, the purpose of LPile analysis was to

determine internal shear and bending moments for selection of an adequate pile section. Tetra

Tech also requested p-y curve definitions for modelling of the H-Piles in structural analysis

software. Modelling of the free head condition for the HP360x132 piles for integral abutmentconsisted of predicting soil resistance over the pile length and internal shear and bending

moments within the pile in response to specified lateral pile head displacements of up to 30

mm and zero moment applied at the pile head. Modelling of the fixed head condition for the

HP360x132 piles for integral abutment consisted of predicting soil resistance over the pile

length and internal shear and bending moments within the pile in response to specified lateral

pile head displacements of up to 30 mm while maintaining zero rotation of the pile head.

Top of pile and final grade configurations used to define the LPile models are summarized in

Table 8-3 and were developed from the Preliminary Drawing for Option 1 in Appendix A.

Although piles are embedded into pile caps, the top of pile elevations were taken as the

underside of pile cap such that top of pile lateral deflections correspond to lateral movement

at the underside of the pile cap. A modulus of elasticity (E) of 200 GPa was used to define the

stiffness of steel piles.

Table 8-3: Top of Pile and Grade Elevation LPile Configurations

Pile Type Pile LocationElevation

Top of PileElevation

Top of Final Grade

610x12.7 Pipe

(Concrete Filled)Pier 232.9 m 231.9 m

HP360x132

(weak axis)

Integral abutment

(annular void to 3 m

below top of pile)

232.4 m1 234.4 m

1. LPile does not support pile heads below existing grade. In this regard, a top of pile elevation of 234.4 m was

used in LPile, and horizontal movement at elevation 232.4 m was estimated from the lateral deflection versus

depth plot at a depth of 2.0 m.

Graphical results summarizing lateral deflection of the pile top (bottom graph) and maximum

bending moment in the pile (top graph) as a function of applied static lateral load at the pile

head and zero applied moment are presented in Figure 3 for the concrete filled pipe piles, and




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Figure 4 for the H-piles. These charts can be used as design charts to estimate lateral

deflection at the underside of pile cap. At the request of Tetra-Tech, p-y curves and additional

graphical results output by LPile for the HP360x132 piles are included in Appendix E, and

summarize the following:

Page 1: p-y curve definitions

Page 2, Top Left Figure: deflection versus pile length

Page 2, Top Right Figure: bending moment versus pile length

Page 2, Middle Left Figure: shear force versus pile length

Page 2, Bottom Left Figure: mobilized pile stiffness versus pile length

Page 2, Bottom Right Figure: load intensity p versus lateral deflection

Amec Foster Wheeler recommends that all piles required to provide lateral resistance should

be embedded a minimum of 2 m into underlying till (i.e. elevation 227.0 m assuming till at

229.0 m) in order to maintain pile toe fixity. This minimum embedment depth shall be included

with the blow count criteria in developing pile acceptance criteria for pile required to provide

lateral resistance. Where this minimum embedment depth into till is not achieved, thenevaluation of pile specific monitoring logs should be undertaken to evaluate the lateral pile

resistance achieved and the impact on foundation performance. This may include drilling of a

test holes next to the pile(s) to confirm soil conditions at the afflicted pile location(s).

8.2.4.4 Inclined (Battered) Piles

Where fill depths are limited to less than 2 m and final design grades remain sufficiently similar

to existing grades, such as is anticipated for this project, significant soil settlement relative to

pile movement is not anticipated at the abutments. In this regard, it is anticipated inclined piles

will be acceptable. Inclined piles may be designed using the recommendations outlined in

Section 8.2.1 to determine the axial capacity of the pile, and horizontal and vertical capacity

determined from the angle of inclination. Inclined piles should be sloped no shallower than

1H:4V.

8.2.5 Minimum Embedment Depth

The minimum required embedment depth of a pile shall be taken as the greater of the

following, as determined in accordance with the recommendation outlined in the previous

sections of this report:

The embedment depth required to provide the required compressive (downwardacting) load resistance (See Section 8.2.1).

The embedment depth required to provide the required tensile (uplift) load resistance,

either due to foundation loads or due to adfreeze forces and/or frost heave pressure

on the underside of foundation elements (See Sections 8.2.3 and 11.2).




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The embedment depth required to provide the required fixity for lateral loadresistance, inclusive of any additional embedment depth required to extend below

the depth of scouring (See Section 8.2.4).

Notwithstanding the minimum embedment depths required to attain the design compressive,

tensile, and lateral resistance requirements, individual pile capacities should be confirmed

during driving based on either WEAP correlations between capacity and blow count/driving

energy, or through PDA monitoring and CAPWAP interpretations.

8.3 Pile Group Effects

Based on review of the preliminary design bridge profiles provided by Tetra Tech (see

Appendix A), Amec Foster Wheeler understood abutments and piers (if required) will be

supported on two rows of driven steel piles.

Generally, piles will behave individually in compression (i.e. Group efficiency η = 1.0) when a

minimum centre-to-centre spacing of fives pile diameters is provided between adjacent piles,

and will behave individually laterally when the center-to-center spacing is greater than five pile

diameters in either the direction transverse to loading (side-by-side), or the direction parallel

to loading (in-line). However, for circumstances in which the final pile layout places piles closer

than the spacing outlined above, interaction between the piles could occur and should be

reviewed by Amec Foster Wheeler during detailed design.

Possible interactive effects include changes to the efficiency of individual piles forming the

pile group, as well as stress overlap. In addition, construction related issues such as possible

heave of piles during driving of adjacent piles become a factor with closely spaced piles, and

needs to be verified with suitable quality control during construction.

8.4 Artesian Groundwater Impact on Semi-Integral Abutments

Amec Foster Wheeler understands that annular voids will be provided around integral

abutment piles to mitigate stresses within the abutments in response to thermal expansion

and contraction of the bridge superstructure. It is further understood that this voids will extend

approximately 3 m below the underside of pile cap, where the underside of pile cap elevation

is approximately 232.9 m for the integral Bridge Option 1. This places the base of the void at

approximate elevation 229.9 m, well with the range of potential development of perpetual

groundwater flow along the piles due to artesian conditions within the underlying till and

bedrock. The potential for developing flowing groundwater conditions through the proposed

annular voids can be mitigated by placing a seal at the base void that will resist artesian

groundwater pressure acting on the underside of the seal. The seal material itself also has to

be resistant to piping.

Discussions during preliminary design meetings with Tetra Tech identified a concrete seal

bonded to the surface of the pile as well as the internal wall of the void medium, likely to

consist of CSP. The concrete seal should be designed to be a minimum of 600 mm thick, oradditional thickness as required such that the sum of the weight of the concrete seal plus the




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bond stress between the seal and the steel pile is equal to 1.4 times the artesian pressure

acting on the underside of the tremie seal. The assumed weight of the concrete may be taken

as 22 kN/m3, and the bond stress between the concrete and steel pile may be taken as

250 kPa. The pressure acting on the underside of the seal should be determined in

accordance with Section 7.0.

Tension in the concrete seal due to bending moments induced by artesian pressure acting on

the underside of the seal should be checked by the structural engineer. The pile and annular

sleeve should be treated as points of support for the tremie seal, and the tremie seal should

be treated as an unreinforced concrete beam.

In accordance with Section 6.11.4.10 of the CHBDC, the following two limit states shall be

considered when settlement of the surrounding ground occurs relative to a pile:

a) ULS of the pile at the neutral plane; and

b) SLS deformations at the pile top

Further in accordance with Article 6.11.4.10 of the CHBDC, unfactored permanent loads

associated with life cycle settlement of the surrounding ground shall be used when predicting

the neutral plane location. Transient loads shall not be included in the prediction of the location

of the neutral plane or settlement.

For clarity, the term ‘downdrag load’ as it is used in the CHBDC is defined in this report as the

drag load transferred to foundation elements by the downward movement of the soil relative

to the foundation element (i.e. pile). It is the integration of negative shaft friction transferred to

the foundation over the zone where soil moves down relative to the foundation element.

Downdrag load does not reduce the geotechnical resistance of a pile; but it must be

considered in combination with structural loads in order to verify that the stress in the pile at

the neutral plane (i.e. the point where the stress in the pile is greatest) does not jeopardize

pile integrity.

With respect to driven steel piles for the bridge structure at this Site, steel piles driven to

practical refusal within the very dense underlying till or on bedrock will develop the majority of

their resistance (i.e. greater than 90 percent) in shaft friction through the till combined with

end-bearing at the pile toe. Final design grades should be reviewed for fill thickness; however,

where fills do not raise final grade above elevation 237 m, it is anticipated that the surcharge

loading due to fill placement will not induce significant settlement within the underlying till. In

this regard, steel piles driven to practical refusal in within the underlying till will not be

susceptible to downdrag (i.e. settlement). Rather, the piles will be subjected to drag load

induced by negative shaft friction over the length of pile in contact with the clay overburden

and overlying embankment fill, where present.

Conservatively, the underside of the drag load zone may be assumed at elevation 229.0 m

(i.e. near the top of the till), and the drag load can be determined using a negative unit shaft

friction of 50 kPa integrated over the length of pile in contact with soil above this elevation.




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For pipe piles, negative shaft friction should only be applied over the exterior surface area of

the pile in contact with the soil. In the case of steel H piles, the surface area should include

the exterior sides of the two flanges plus twice the depth of the web.

10.1.1 Earth Pressure Coefficients

The determination of lateral earth pressures will be required for the design of abutment

wingwalls and other substructures. Table 10-1 provides recommended earth pressure

coefficients for the active, passive and “at rest” earth pressure cases, and total unit weights

for various soil backfill types assuming horizontal grades and a vertical wall. The earth

pressure coefficients should be reviewed during detailed design for sloping grades and wall

faces.

Table 10-1: Earth Pressure Coefficients and Soil Unit Weights

Soil Type

Active

Pressure

Coefficient

Ka

“At Rest”

Pressure

Coefficient

Ko

Passive

Pressure

Coefficient

Kp

Total

Soil

Unit

Weight

(kN/m3)

Friction

Angle

(deg)

Between

Soil and

Concrete

Granular

Fill

Well Compacted (35°) 0.27 0.431 2.46 22 23

Moderately Compacted (30°) 0.33 0.501 2.00 21 20

Cohesive

Fill

Well Compacted (20°) 0.49 0.661 1.36 18 16

Moderately Compacted (15°) 0.59 0.741 1.13 17 12

1. In the case of unyielding walls exposed to frost penetration above the groundwater table, it is recommended that Ko = 1.0,

be used to account for lateral frost pressures1

The passive earth pressure coefficients provided in Table 10-1 include a reduction factor of

1.5 to account for the partial mobilization of passive resistance that is consistent with the smallwall displacements expected under operational conditions. Relatively large wall

displacements would be necessary to realize full passive resistances. To determine the

factored resistance, the resistance factor (Φ) of 0.5 should be applied to the passive earth

pressure.

Where sub-drainage will not be provided behind a wall, buoyant soil unit weights should be

used, and a hydrostatic pressure component will need to be included in the design. Buoyant

soil unit weights are determined by subtracting the unit weight of water (10 kN/m3) from the

given total unit weights. The recommended design groundwater level may be taken as

elevation 235 m.

1 As per Canadian Foundation Engineering Manual, 3rd Edition, P. 429, an earth pressure coefficient K=1 should be used incombination with insulation for highly frost susceptible soils.




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30 September 2016


The “at rest” (Ko) earth pressure should be used in the case of unyielding walls. To attain

active earth pressure (Ka) conditions, the displacement at the top of a cantilevered wall should

be at least 0.01 times the height of the wall. In the case of unyielding walls exposed to frost

penetration above the groundwater table, it is recommended that Ko = 1.0, be used to account

for lateral frost pressures. However, where lateral frost pressures need to be considered in

the design, the site-specific configurations of the walls or sub-structures should be reviewed

by qualified geotechnical personnel to explore alternatives in reducing the frost pressures

10.1.2 Calculation of Earth Pressure Distribution and Surcharge Loads

The magnitude and distribution of the lateral earth pressures on below grade structures will

depend on such factors as the rigidity of the below grade structure; the degree of compaction

of the backfill against the structure; the backfill soil type; the slope angle at the structure/soil

interface; and the subsurface drainage and groundwater conditions over the height of the

structure. It is anticipated that a sloped excavation will be implemented for construction of

below grade foundation structures, which will necessitate the placement of backfill behindbelow grade structure walls. The magnitude and distribution of the lateral earth pressures (P)

on below grade structures will depend on the degree of compaction of the backfill. In addition

to earth pressures, lateral stresses generated by any applicable surcharge loads also need to

be evaluated in the design. Recommended earth pressure distributions for light to moderate

and moderate to well compacted backfill cases, as well as for line or point surcharge loads,

are discussed in the following sections. The recommended earth pressure distributions are

for preliminary design only, and their applicability should be reviewed during detailed design.

10.1.2.1 Moderate to Well Compacted Backfill Case

Where subgrade support on the surface of the retained soil behind a wall is required, the

backfill against the wall will need to be compacted to at least 95 percent Standard Proctor

maximum dry density. In this case, the design earth pressure distribution should adopt a

combined trapezoidal/triangular distribution as shown in Figure 5 attached to account for the

induced lateral pressures due to compaction. Figure 5 also provides the relationships to be

used in the calculation of the compaction induced earth pressures, and tabulated loads (P)

generated by typical compactors. The earth pressure coefficients to be used in the calculation

of the lateral pressures should be those applicable to the backfill types given in Table 10-1.

10.1.2.2 Surcharge Loads

In addition to earth pressures, lateral stresses generated by surcharge loads, such as point

loads from heavy trucks, also need to be evaluated in the design. For line or point surcharge

loads, the lateral pressures should be determined using the relationships given in Figure 6. Inthe case of uniformly distributed surcharge loads, such as those due to the fluid contents

beneath a tank base (for design of the concrete ring beam), or those acting on the surface of

the retained soil, the induced lateral earth pressure may be determined by multiplying the

surcharge load by the appropriate earth pressure coefficient.




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10.1.3 Load Factors

For the Limit States Design procedure for walls, the following Load Factors should be applied

to loads calculated from the pressure distributions given above.

For earth loads acting on walls, a Load Factor of 1.25 is recommended for sustainedloads.

For hydrostatic loads acting on walls, a Load Factor of 1.1 is recommended.

For live surcharge loads acting on walls, the Load Factor of 1.5 should be used.

The above load factors should be applied to loads lead to instability of the walls.

11.1 Frost Penetration Depth

The upper stratigraphy at the Site is considered moderately to highly frost susceptible in the

presence of a free supply of water, and as such, frost effects should be considered for

foundations or surface structures sensitive to movement. Based on historical temperature

data for the area, a design frost penetration of 2.4 m below final grade is recommended in

areas that will not have regular snow or vegetative ground cover. It should be noted that this

recommended frost penetration depth extends both vertically and laterally behind final surface

(i.e. extends 2.4 m behind the headwall).

11.2 Pile Foundations

Frost forces applied to pile foundations include adfreeze pressures acting along the pile shafts

within the depth of frost penetration. If pile caps are used and extend beyond the perimeter of

the underlying pile, then frost heave forces acting on the undersides of the pile caps, as well

as any connecting supports (i.e. lateral tie between the piles) will also need to be considered.

11.2.1 Frost Heave

To reduce the potential of frost heave pressures, a void-forming product should be installed

beneath the underside of the pile caps and any other structural element located within the

depth of frost penetration above the groundwater table. The recommended minimum

thickness of the void should be 150 mm. Alternatively, a compressible material may be used

in lieu of a void forming material, and the uplift pressures may be taken as the crushing

strength of the compressible medium. It is recommended that a frost heave of 150 mm be

assumed in determining the required thickness for the void-filler and the associated uplift

pressures associated with the thickness used.

The finished grade adjacent to each pile cap should be capped with well compacted clay and

sloped away so that the surface runoff is not allowed to infiltrate and collect in the void space

or saturate the compressible medium. If the soil layer within which the underside of the pile

cap is located is free draining (i.e. sand or gravel) such that infiltration cannot accumulate




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30 September 2016


within a void or saturation of the compressible medium, then a clay cap at finished grade may

not be required.

The use of void-forming product below the groundwater is unfeasible. In instances where

groundwater is located within the recommended depth of frost penetration, the underside of

foundation elements such as pile caps should extend below the depth of frost penetration to

mitigate frost heave development on the underside of the foundation element.

11.2.2 Adfreeze Stresses

Resistance to adfreeze and frost heave forces will be provided by the sustained vertical loads

on the foundation, the buoyant weight of the foundation and dead weight of the structure, and

the soil uplift resistance component provided by the length of the piles extending below the

depth of frost penetration. In the case of driven steel piles, the adfreeze force acting on the

pile may be determined assuming an unfactored unit adfreeze stress of 65 kPa applied to the

exterior surface of the pile and supported foundation elements (i.e. pile caps) located within

the zone of frost penetration. A load factor of 1.25 should be applied to obtain the factoredadfreeze stress.

Adfreeze stresses along the sides of pile caps and buried substructures can be reduced by

the installation of a ‘bond-break’ or ‘friction reducer’ within the zone of frost penetration.

Friction reducers could consist of a system of poly wrapped sono-tubes. A smooth

geosynthetic liner material, fixed to the shaft of the pile or to the sides of the pile cap would

also be a suitable bond-break.

In the case of straight shaft piles supporting lightly loaded unheated facilities, the piles should

be embedded a minimum of 7 m below final grade in order to provide sufficient frictional

resistance against potential uplift due to adfreeze stresses.

12.1 Design Philosophy

Slope stability analyses were completed for the proposed 4H:1V headslope configuration for

Bridge Option 1 and the proposed 5H:1V headslope configuration for Bridge Option 2. Slope

stability analyses were also completed for three cross-sections (Cross Section A, B, C)

selected by Tetra-Tech through the proposed grouted slope protection to be constructed

between the proposed bridge alignment and the existing railway. Specifically, slope stability

analyses of the headslopes and grouted slope protection were completed to determine the

Factor of Safety (FS) against rotational failure to satisfy minimum target factor of safety levels

for both long term conditions (i.e., when all construction induced pore water pressures are

fully dissipated), and short term conditions (i.e. a dewatered creek as is expected forconstruction of slope protection works). Minimum factor of safety targets are discussed in

Section 12.1.1.




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12.1.1 Design Criteria

Limit equilibrium stability analyses were performed to confirm that; in keeping with the

standard local practice and city of Winnipeg Water Way permit requirements; the following

minimum target factor of safety would be achieved:

A minimum factor of safety of 1.5 for long term stability under ‘normal’ seasonalgroundwater and creek levels; and

A minimum factor of safety of 1.3 for short term stability under ‘short term’ conditions.Examples of short term conditions include spring flooding followed by rapid drawdown

to normal summer creek level, and temporary dewatering of the creek for construction.

Creek levels and groundwater conditions are discussed in Section 12.1.4.

12.1.2 Methodology and Model Geometry

All slope stability analyses were conducted using SLOPE/W, a limit equilibrium software

package developed by Geo-Slope International. Slope stability models for each of the two

proposed headslope configurations and each of the three cross-sections taken through the

grouted slope protection are shown in Appendix F. The models were developed by Amec

Foster Wheeler from the elevations and the bridge and slope configurations shown in the

preliminary plans provided by Tetra Tech, included in Appendix A.

12.1.3 Soil Stratigraphy and Soil parameters

Soil conditions along the project site were discussed in Section 4.0. In summary, Amec Foster

Wheeler idealized the stratigraphy into individual layers by soil type, as illustrated in the slope

stability outputs included in Appendix F. Table 12-1 summarizes the material strengthparameters and unit weights assigned for each of the soil types/layers. The parameters

presented in Table 12-1are based on Amec Foster Wheeler’s previous experience with similar

soils in the Winnipeg area. Other than erosion and scour failures south of the existing bridge

structure, there were limited signs of any slope movement. In this regard, post peak shear

strengths have been assumed for all clay soils, which is in keeping with accepted practice

The Morgenstern-Price method with a half sine variation of inter-slice forces was used for all

analyses.




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30 September 2016


Table 12-1: Summary of Slope Stability Material Parameters

LayerUnit Weight,

γ (kN/m3)

Pore Pressure

Coefficient ru

Effective Stress Parameters

Cohesion, c’

(kPa)

Friction Angle,

ϕ’ (°)

Clay Fill 17.5 NA 2 20

Native Brown Clay 17.5 NA 5 15

Native Grey Clay 17.5 NA 5 15

Glacial Silt Till 20 NA 1 35

Rip-Rap (Unbound

Aggregate)20 NA 0 35

Grouted Slope

Protection (Cracked)23 NA 0 35

12.1.4 Piezometric Conditions and Creek Levels

Groundwater conditions observed at the test hole locations is discussed in Section 4.3. Based

on the vibrating wire measurements obtained at the site, groundwater levels within the till and

clay overburden are considered to be governed by artesian pressure within the underlying

bedrock. In summary, porewater pressures were modelled using a phreatic surface; applied

to each soil layers; defined using Cartesian coordinates. Based on the VW measurements, a

phreatic surface of 234.4 m was assigned to the till stratum, and was assumed constant

throughout the year. Porewater pressure with the clay and clay fill were modelled based on

the VW measurement, and were defined using a minimum phreatic surface elevation taken

as equal to the water level in the creek, and phreatic surface elevations above the creek level

taken as the lower elevation given by the following:

Elevation 233.8 m, or

Underside of rip-rap or grouted slope protection.

Porewater pressure within the clay fill and clay established from the summer VW readings

were conservatively assumed to remain constant throughout the year. Notwithstanding,

groundwater monitoring data should continue to be collected and reviewed during the detailed

design phase to confirm that assumed groundwater conditions.

With respect to water levels within the Creek, the preliminary design drawings provided byTetra Tech (Appendix A) outlined a Q50% (i.e. 1 in 2 year) water level of 233.370, and a Q1%

(i.e. 1 in 100 year) water level of 234.520. The existing water level surveyed on 22 June 2016

was approximately 232.67 m. The Q50% level defines the Creek levels for the ‘Normal’

groundwater and creek configuration; while the Q1% defines the creek level for the 1in 100

year flood condition. A dewatered creek defines the ‘Extreme’ model condition.




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30 September 2016


12.2 Slope Stability Results

Table 12-2 presents the resulting Factor of Safety (FS’s) results for ‘normal’ for long term

stability of the headslopes and cross-sections taken through the proposed grouted slope

protection discussed in Section 12.1.2.

Table 12-2: Summary of Slope Stability Results for Overburden Phreatic Surface

Elevation 233.8 m, and Artesian Total Head of 234.4 m in the underlying Till

Model and Slope

Configuration

General Location of

Potential Slip

Surface (PSS)

Critical Factor of Safety (FOS) for Various Creek Levels

Empty

Channel

232.67

(22 June 2016)

233.37

(Q50)

234.52

(Q1)

(Extreme

Condition)

(Existing

Condition)

(Normal

Condition)

(Flood

Condition)

Bridge Option 1 –

5H:1V Headslope

Configuration

All PSS 1.31 1.69 1.87 2.20

Bridge Option 2 –

4H:1V Headslope

Configuration

All PSS 1.11 1.45 1.62 1.93

Grouted Slope

Protection – Section AAll PSS 1.01 1.90 NC NC

Grouted Slope

Protection – Section B

PSS originating at

top of railway –

Composite Slope

1.13 1.22 1.34 1.44

PSS originating at

crest of rip-rap /

base of rail

embankment

1.36 1.54 1.81 2.18

Grouted Slope

Protection – Section C

PSS originating at

top of railway –

Composite Slope

1.06 1.13 1.20 1.28

PSS originating at

crest of rip-rap /

base of rail

embankment

1.16 1.32 1.47 1.75

Design Criteria (Minimum FOS Requirement) 1.3 N/A 1.5 1.3

NC = not completed. FOS result will be greater than 1.5 by inspection.

12.2.1 Headslope Stability

With respect to Bridge Option 1 and a 5H:1V headslope configuration, slope stability results

outlined in Table 12-2 indicate that a 5H:1V headslope configuration will achieve the design

criteria and minimum factor of safety requirements set forth in Section 12.1.1 withoutrequirement for additional slope improvement (or stabilization) measures; provided that the




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30 September 2016


headslopes and full width of the channel bottom are protected with 675 mm thick rip-rap

erosion protection.

With respect Bridge Option 2 and a 4H:1V headslope configuration, slope stability results

outlined in Table 12-2 indicate that the factor of safety of a 4H:1V headslope configuration is

highly susceptible to fluctuation in the water level of Sturgeon Creek. In fact, factor of safety

results are less than 1.5 for the water level surveyed on the 22 June 2016. In this regard, use

of a 4H:1V headslope configuration could necessitate additional slope improvement (or

stabilization) measures. Shear keys and/or stone columns comprise the slope improvement

measures most commonly employed in Winnipeg. Critical slip surfaces illustrated within the

slope stability outputs in Appendix F are deep seated and extend along the top of the till. In

this regard, shear keys and stone columns would have to extend to slightly below the surface

of the till in order to intersect potential slip surface along the surface of the till and achieve

minimum factors of safety targets. However, given artesian conditions within the underlying

till and the associated risk of developing a flowing artesian conditions and/or basal instability(See Section 6.3), the use of shear keys and/or stone columns is not recommended at this

site. If required, alternative slope improvement measures that could be considered through

detailed design include anchored sheet pile walls embedded to toe fixity in the underlying till.

Based on the discussion above, and giving consideration to both design and construction

costs as well as potential construction risks associated with artesian conditions within the

underlying till, unreinforced headslopes steeper than 5H:1V are not recommended.

12.2.2 Grouted Rip Rap Slopes

Due to the complexity of transitioning slopes over short distances as confined to the space

between the existing railway embankment and the proposed bridge structure, 3D modelling

of the stability of slopes between the structures is recommended for detailed design. 2D slope

stability analyses are not considered suitably representative of the real interaction between

transitioning slopes. Notwithstanding, 2D slope stability was undertaken to qualitatively

assess the range within which 3D slope stability results could be expected to fall.

Two dimensional slope stability analyses were undertaken on three cross-sections taken

through the proposed grouted rip-rap slopes. The proposed rip-rap configuration and the

locations of the cross-sections are shown on Tetra-Tech drawing SK3 included in Appendix

A. It should be noted that cross-sections B and C are composite sections, whereby the bottom

of the sections originate at the creek; the sections extend perpendicular with the slope to the

toe of the existing railway embankment (i.e. above the top of the proposed rip-rap and

approximate invert of the existing ditch running parallel with the railway); and then the sectionschange direction to extend up to the top of the existing railway embankment.

Slope stability analysis results are summarized in Table 12-2 for various Sturgeon Creek water

elevations. In summary, slope stability analyses indicate slip surfaces originating at the crest

of rip-rap / base of rail embankment (i.e. prior to the inflection in the composite section), are

stable under normal groundwater conditions and flood conditions. These same slip surfaces

are inherently unstable for the extreme dewatered creek condition expected to only occur




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30 September 2016


during construction. In this regard, slope stabilization measures will be required to maintain

slope stability during construction. It is envisaged analyses for detailed design will explore

anchored sheet piling installed at the toe of the slopes (i.e. at the crest of the creek) to fixity

in the underlying till to intersect slip surfaces extending into the creek bed.

Notwithstanding the stability of slip surface originating at the crest of rip-rap / base of rail

embankment, the slope stability result in Table 12-2 for composite Section B does not meet

the target FOS for the ‘normal’ creek level. Furthermore, the slope stability results in Table

12-2 for composite Section C do not meet the target FOS values for either the ‘normal’ or

‘flood’ creek levels. Notwithstanding, 2D representation of a three dimensional cross-sections

is not an accurate representation of real stress conditions. Due to the complexity of

transitioning slopes over short distances as confined to the space between the existing railway

embankment and the proposed bridge structure, 3D slope stability modelling is recommended

for detailed design. It is anticipated that 3D slope modelling, combined with implementation

of anchored sheet piling at the toe of the slopes to support temporary dewatering of the creek,will support the proposed Grouted Rip Rap Slope configuration.

12.2.3 Cofferdam and Creek Dewatering (Short Term Stability)

Amec Foster Wheeler understood Tetra-Tech proposes to dewater the creek (using coffer

dams) in order to evaluate the existing structure between the existing railway and the existing

Saskatchewan Avenue Bridge; as well as to undertake slope works required for the final

bridge design. It should be noted that dewatering of the creek, and in particular removing soil

in the creek, will negatively impact slope stability and potentially jeopardize the existing railway

embankment.

The factor of safety of the proposed headslopes and proposed rip-rap configuration for a

dewatered creek condition; assuming no further subexcavation of the creek bottom or slopes;

may be taken as the results outlined in Table 12-2 for the “Empty Channel” condition. In

summary, slope stability results indicate that sideslopes steeper than 5H:1V, including the

grouted rip-rap slopes, will all fail to meet the minimum target FOS of 1.3 for an ‘Extreme’

Condition given by a dewatered creek condition. In this regard, slope stabilization measures

will be required to meet the minimum target factor of safety for slope stability during

construction. It is envisaged analyses for detailed design will explore anchored sheet piling

installed at the toe of the slopes (i.e. at the crest of the creek) to fixity in the underlying till to

intersect slip surfaces extending into the creek bed.

Slope stability analyses to satisfy short term construction activities (i.e. dewatering of the creek

and subexcavation of the creek bed) shall by undertaken during detailed design once the finalbridge alternative and temporary construction requirements have been determined. Due to

the complexity of transitioning slopes over short distances as confined to the space between

the existing railway embankment and the proposed bridge structure, 3D slope stability

modelling is recommended for detailed design.




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12.3 Slope Stability Conclusions and Recommendations for Detailed Design

Based on the slope stability analyses and results discussed in Sections 12.1 and 12.2,

conclusions and recommendations for detailed design are as follows:

The stability of both existing slopes and proposed design slopes is very sensitive tochanges in the water level of Sturgeon Creek.

A 5H:1V headslope configuration will achieve the design criteria and minimum factorof safety requirements set forth in Section 12.1.1 for all potential creek levels,

including a dewatered creek condition (neglecting basal stability issues due to

artesian pressure).

Although 4H:1V headslopes are marginally stable for the ‘normal’ condition definedby a Creek Level of 233.37 m, the Factor of Safety for the existing creek level of

232.67 m at the time of this investigation falls below the minimum target of 1.5 for the

normal condition. Based on interpolation of the FOS results for Creek Levels 232.67

and 233.37, a minimum creek level of about 232.9 m would need to be assured to

meet the minimum target FOS of 1.5 for ‘Normal’ Conditions. Furthermore, 4H:1V

headslopes do not achieve the minimum target FOS of 1.3 for the dewatered creek

condition.

Giving consideration to bullets 2 and 3 above, as well as consideration to design andconstruction costs as well as potential construction risks associated with artesian

conditions within the underlying till, Amec Foster Wheeler recommends a maximum

(i.e. steepest) headslope configuration of 5H:1V.

Excluding concerns related to basal heave discussed in Sections 6.3 and 7.0,dewatering of Sturgeon Creek will temporarily reduce the stability of both existing

slopes and proposed construction slopes to less than the minimum recommended

target of 1.3. In this regard, slope stabilization of will be required for the construction

condition (dewatered creek).

It is envisaged slope stabilization considerations for detailed design will exploreanchored sheet piling installed at the toe of the slopes (i.e. at the crest of the creek)

in order to intersect potential slip surface extending into the creek bed. Specifically,

the sheet piles will be embedded to fixity in the underlying till, and if required,

anchored near the top in order to mitigate lateral deflection of the sheet pile. Due to

artesian pressure within the underlying till and the risk of basal heave and

development of a flowing artesian condition, construction of conventional shear keys

or rock filled caissons to stabilize slopes is not recommended unless

depressurization of the aquifer were undertaken.

Due to the complexity of transitioning slopes over short distances as confined to the

space between the existing railway embankment and the proposed bridge structure,

3D slope stability modelling is recommended for detailed design of the grouted rip-rap configuration. It is anticipated that 3D slope modelling, combined with

implementation of anchored sheet piling at the toe of the slopes to support temporary




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30 September 2016


dewatering of the creek, will support the proposed Grouted Rip Rap Slope

configuration.

Where concrete elements outlined in this report and all other concrete in contact with the local

soil will be subjected in service to weathering, sulphate attack, a corrosive environment, or

saturated conditions, the concrete should be designed, specified, and constructed in

accordance with concrete exposure classifications outlined in the latest edition of CSA

standard A23.1, Concrete Materials and Methods of Concrete Construction. In addition, all

concrete must be supplied in accordance with current Manitoba and National Building Code

requirements.

Based on significant data gathered through previous work in southern Manitoba, water soluble

sulphate concentrations in the soil are typically in the range of 0.2% to 2.0%. As such, the

degree of sulphate exposure at the site may be considered as ‘severe’ in accordance with

current CSA standards, and the use of sulphate resistance cement (Type HS or HSb) isrecommended for concrete in contact with the local soil. Furthermore, air entrainment should

be incorporated into any concrete elements that are exposed to freeze-thaw to enhance its

durability.

It should be recognized that there may be structural and other considerations, which may

necessitate additional requirements for subsurface concrete mix design.

14.1 Pavement Design Methodology

Asphalt Concrete Pavement designs for Manitoba Avenue have been developed using the

1993 AASHTO Guide for Design of Pavement Structures.

14.2 Design Vehicle and Traffic

Based on discussions with Tetra Tech, Amec Foster Wheeler understood that pavement

design were to be developed assuming an initial two-way average annual daily traffic (AADT)

of 4700 vehicles per day. In terms of reducing the Two-Way AADT to a single lane design

volume, further discussions with Tetra Tech indicated a direction distribution factor of 0.5 and

a lane distribution factor of 1.0.

In terms of design vehicles to determine pavement loading, discussions with Tetra Tech

indicated truck volumes shall be taken as 2.5% of the AADT, and that the design truck shall

consist of a 5-axle single steer tandem truck tractor/tandem trailer combination (or 3S2 truck)

with a gross vehicle weight of about 40,000 kg (88,185 lbs). Using axle load factors from

AASHTO, such a truck has a single pass ESAL equivalency of about 3.56 ESALs per pass

for ACP. The remaining 97.5% of traffic volume was assumed to be passenger vehicles with

a gross vehicle weight of about 1,800 kg (4,000 lbs).




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14.3 Subgrade Resilient Modulus

Pavement sections have been designed using a nominal effective subgrade resilient modulus

(Mr) of about 30 MPa (4500 psi), or an approximately equivalent California Bearing Ratio of

3.0 percent. The above subgrade resilient modulus has been determined for the soil

conditions observed at the test hole locations, and has been calculated accounting for the

climatic region and associated seasonal variations in subgrade strength using the

serviceability method outlined in Section 2.3.1 of the 1993 AASHTO Guide for Design of

Pavement Structures.

14.4 Granular Base Course and Subbase Course Materials

Table 14-1 summarizes key parameters for granular base course and granular subbase

course materials used in pavement design. Granular base course and granular subbase

course materials were selected in accordance with locally available aggregates. Resilient

moduli as shown were selected based on correlation with layer coefficients commonly used

in local practice and design of pavements using the 1993 AASHTO Guide for Design ofPavement Structures. Anticipated maximum dry unit weights and optimum moisture contents

have been specified based on Amec Foster Wheeler Winnipeg’s database of maximum dry

density results.

Table 14-1: Summary of Granular Pavement Structure Materials

Layer Material Resilient Modulus(MPa / psi)

Maximum DryUnit Weight(kg/m3 / pcf)

Optimum MoistureContent (%)

Granular BaseCourse &

Drainable StableBase

20 mm minuscrushed limestone 207 / 30,000 2,200 / 137 8.3

Granular SubbaseCourse Layer 1

50 mm maxcrushed limestone 172 / 25,000 2,300 / 144 6.3

Granular SubbaseCourse Layer 1

150 mm minuslimestone 172 / 25,000 2,000 / 125 6.3

14.5 Subgrade Preparation

Typical construction recommendations for subgrade preparation and pavement construction

provided locally are as follows:

1. Excavate to design subgrade elevation, which should be taken as the top of

the pavement minus the pavement thickness and the recommended

minimum base course and subbase gravel structure for the specified

pavement structure. Further excavation should be conducted as required to

remove organic or otherwise unsuitable soils.




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2. Stripping and excavation to subgrade design elevation should be completed

in such a manner as to minimize disturbance of the subgrade. In this regard,

Amec Foster Wheeler recommends that excavation be completed using a

backhoe equipped with a smooth bladed bucket operating from the edge of

the excavation. Further, no construction equipment should be allowed on the

exposed subgrade until an assessment of the subgrade has been completed

by knowledgeable and experienced geotechnical personnel.

3. Once the final subgrade elevation has been achieved, an assessment of the

subgrade shall be completed in order to identify any localized loose, ‘weak’,

or soft areas prior to trafficking the subgrade and/or prior to fill operations.

Ground conditions permitting, assessment of the subgrade should consist of

proof-rolling the subgrade with multiple passes of a fully loaded tandem.

Notwithstanding, the ability of a subgrade to support proof-roll loads is

subject to change throughout construction as a result of changing moistureconditions, and in this regard, proof-rolling may not be possible. The exposed

subgrade and feasibility of proof-rolling of the subgrade should be visually

evaluated by qualified geotechnical personnel throughout stripping and

subgrade preparation operations.

4. Loose, ‘weak’, or soft areas identified either visually or by proof-rolling should

be sub-excavated below design subgrade as required to achieve a

competent subgrade stratum up to a maximum of 400 mm below grade, and

replaced with engineered fill material, as directed by the engineer at the time

of construction. Where silt remains at the subgrade elevation, specific

backfilling methods and procedures and use of select fill materials (such as

100 mm down crushed limestone underlain by a geotextile separator) may

be required.

5. Protect the exposed subgrade from frost, desiccation (drying), and

inundation (wetting) both during and following construction. To reduce

accumulation of surface runoff and softening of the subgrade, rough grades

should be designed to reduce the potential for ponding of water on the

surface and to provide positive drainage towards the perimeter of the

subgrade area and/or collection areas as quickly as possible, both during

and following subgrade preparation.

6. Depending on disturbance and protection of the subgrade, exposed

subgrades that are highly disturbed (i.e. rutted), or desiccated or inundated

outside of the acceptable range of the optimum moisture content (i.e. more

than 3 percent below or 3 percent above OMC), should be re-conditioned

and re-compacted prior to fill placement. If required, re-conditioning of the

subgrade should consist of scarifying the subgrade to a minimum of 200 mm

below grade, moisture conditioning dried or wetted soil to between OMC and

3 percent above OMC, and compacted to a minimum of 95 percent of

standard Proctor maximum dry density (SPMDD). If excavation to subgrade




Winnipeg, Manitoba

30 September 2016


minimizes disturbance of the subgrade and the subgrade is stable and within

an acceptable moisture state, then scarification and re-compaction of the

subgrade is not required.

7. Fill materials, if required between the subgrade elevation and the underside

of the pavement structure, should consist of additional granular subbase. The

fill material should be placed in 150 mm thick lifts and uniformly compacted

to 98% of SPMDD. Alternatively, suitable approved clay fill could be used,

provided that it is free of deleterious materials, moisture conditioned to wet

of optimum (preferably two to five percent wet of optimum) and compacted

uniformly to 98 % of SPMDD.

8. The 50 mm minus granular subbase should be placed in maximum 200 mm

thick lifts (or reduced lift thicknesses as governed by the compactive abilities

of the compaction equipment) and uniformly compacted to a minimum of 98

percent of SPMDD at ± 3 percent of OMC to the bottom of the base coursedesign elevation.

9. The granular base course should be placed in maximum 200 mm thick lifts

(or reduced lift thicknesses as governed by the compactive abilities of the

compaction equipment) and uniformly compacted to a minimum 100 percent

of SPMDD at ± 3 percent of OMC to the bottom of the asphalt design

elevation.

10. Qualified geotechnical personnel should monitor the quality and placement

of gravel and the compaction of the gravel should be monitored by field

density testing at regular frequencies. The density of each lift should be

tested to confirm that adequate compaction has been achieved before

placing the next lift.

11. Asphalt should be compacted to a minimum 98 percent of a 75 blow Marshall

Density.

14.6 Asphalt Concrete Pavement (ACP) Alternative

Table 14-2 presents a recommended HMA pavement sections for a Target Reliability level of

90. The first pavement section does not include any consideration to loss of serviceability level

resulting from swelling and freezing of the subgrade soils. The sections have been developed

using the design methodology, traffic loading, and subgrade and base course design

properties outlined in Sections 1.1 through 1.5. Based on discussions with Tetra Tech and

their experience with City of Winnipeg preferences for the Ness Avenue pavement structure,additional assumptions required for selection of other HMA pavement design inputs were as

follows:

Design Life: 25 years

Annual Traffic Growth: 0.5%

Design Serviceability Level: 4.2

Terminal Serviceability Level: 2.5




Winnipeg, Manitoba

30 September 2016


Standard Deviation: 0.45

Structural Layer Coefficient for Asphalt Concrete Pavement: 0.40

Structural Layer Coefficient for Base Course: 0.14

Structural Layer Coefficient for Granular Sub-base: 0.10

Table 14-2: ACP Pavement Alternative for 90% Reliability

Serviceability Loss Due to Swelling and Frost Heave Included? No YesLayer Name Thickness (mm) Thickness (mm)

Surface Asphalt Layer – PG58-40 50 50Asphalt Layer # 2 (Two-Lifts) – PG58-34 90 9520 mm Crushed Limestone Base Course 75 75

50 mm MAX Crushed Limestone Sub-base 150 15050 mm MAX, 100 mm minus, or 150 minus Crushed Sub-base 360 515

Non-Woven Geotextile Fabric (At Subgrade) Yes YesTotal Structure Depth 725 885Structural Number (SN) 120.1 137.7

The estimated loss of serviceability versus pavement age is presented in Figure 7, and has

been estimated based on an estimated swell rate constant of 0.11 for the anticipated subgrade

conditions, and an estimate potential vertical rise of about 46 mm (1.8 inches) over a 4.6 m

thick swelling zone (i.e. given by the depth of the active zone). From the figure, a serviceability

loss of 0.62 is observed over the duration of a 25 year design life. If loss of serviceability due

to swelling and frost heave is not considered over the design life of the pavement, than it isestimated that the terminal serviceability level of 2.5 would occur at about year 14 as opposed

to the 25 year design life. If swelling and frost heave is considered, than the thicker pavement

structure in Table 14-2 would be required to satisfy the target 25 year design life.

All engineering design recommendations presented in this report are based on the

assumption that an adequate level of testing and monitoring will be provided during

construction and that all construction will be carried out by a suitably qualified contractor

experienced in foundation and earthworks construction. An adequate level of testing and

monitoring is considered to be:

for earthworks: full-time monitoring and compaction testing.

for deep foundations: design review and full time monitoring during construction.




Winnipeg, Manitoba

30 September 2016


for concrete construction: testing of plastic and hardened concrete in accordancewith the latest editions of CSA A23.1 and A23.2; and

review of concrete supplier’s mix designs for conformance with prescribed and/or

performance concrete specifications.

Amec Foster Wheeler requests the opportunity to review the design drawings, and the

installation of the foundations, to confirm that the geotechnical recommendations have been

correctly interpreted. Amec Foster Wheeler would be pleased to provide any further

information that may be needed during design and to advise on the geotechnical aspects of

specifications for inclusion in contract documents.

The findings and recommendations presented in this report were based on geotechnical

evaluation of the subsurface conditions and limited groundwater data observed during the site

investigation described in this report and based on the bridge configurations provided to Amec

Foster Wheeler. If conditions other than those reported in this report are noted during

subsequent phases of the project, or if the assumptions stated herein are not in keeping with

the design, this office should be notified immediately in order that the recommendations can

be verified and revised as required. Recommendations presented herein may not be valid if

an adequate level of inspection is not provided during construction, or if relevant building coderequirements are not met.

Soil conditions, by their nature, can be highly variable across a site. The placement of fill and

prior construction activities on a site can contribute to the variability especially in near surface

soil conditions. A contingency should always be included in any construction budget to allow

for the possibility of variation in soil conditions, which may result in modification of the design

and construction procedures.

This report has been prepared for the exclusive use of Tetra Tech WEI Inc., and their agents,

for specific application to the project described in this report. The data and recommendations

provided herein should not be used for any other purpose, or by any other parties, without

review and written advice from Amec Foster Wheeler. Any use that a third party makes of this

report, or any reliance or decisions made based on this report, are the responsibility of those

parties. Amec Foster Wheeler accepts no responsibility for damages suffered by a third party

as a result of decisions made or actions based on this report.

Project File No. WX17918 Amec Foster Wheeler Preliminary Design Geotechnical Report Environment & Infrastructure Saskatchewan Avenue over Sturgeon Creek Culvert Replacement Winnipeg, Manitoba 30 September 2016


FIGURES

TH

04

TH

03

TH

01

TH

05

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06

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01

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02

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03

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400

600

800

1000

1200

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LATERAL RESPONSE OF CONCRETE FILLED DN600 PIPE PILE(12.7 mm WALL)

0.0

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20.0

30.0

40.0

50.0

60.0

70.0

80.0

90.0

100.0

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Applied Lateral Load at Pile Head (kN)

Q=1800 kN, M=0 kN•m, Fixed Head Q=1800 kN, M=0 kN•m, Free Head

TETRA TECH WEI INC.PROJECT:

WX17918

DATE:

AUGUST 2016

DRAWING:

FIGURE 3

SASKATCHEWAN AVENUE OVER STURGEON CREEK LATERAL RESPONSE OF DN600x12.7mm PIPE PILES

CONCRETE FILLED - FIXED & FREE HEAD

0

20

40

60

80

100

120

140

160

180

200M

axim

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om

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)

LATERAL RESPONSE OF HP360x132 INTEGRAL ABUTMENT PILES(Annular void to 3 m below underside of pile cap)

0.0

10.0

20.0

30.0

40.0

50.0

60.0

70.0

80.0

90.0

100.0

0 5 10 15 20 25 30 35

Max

imu

m S

hea

r in

Pile

(kN

)

Lateral Deflection at Underside of Pile Cap (mm)

Q=1630 kN, S= 0, Fixed Head Q=1630 kN, M= 0, Free Head

TETRA TECH WEI INC.PROJECT:

WX17918

DATE:

AUGUST 2016

DRAWING:

FIGURE 4

SASKATCHEWAN AVENUE OVER STURGEON CREEKLATERAL RESPONSE OF HP360x132 INTEGRAL ABUTMENT PILES

BENDING ABOUT WEAK AXIS - FIXED & FREE HEAD

FOR Zc <- Z d-<

FOR Z > d

P (ROLLER LOAD) = DEAD WT. OF ROLLER + CENTRIFUGAL FORCE

WIDTH OF ROLLER

(SEE TEXT OF REPORT)

= SOIL UNIT WEIGHT(SEE TEXT OF REPORT)

EARTH PRESSURE COEFFICIENTS

K = Ko ("AT REST") OR Ka (ACTIVE CASE)

TYPICAL COMPACTOR LOADS (P)

CompactorLOAD (P)

kN/mCompactor

LOAD (P)kN/m

Bowmag TSE 31 Bowmag BW122PD 36

Bowmag 60S 32 Bowmag 142PDB 47

Bowmag 65S 22 Bowmag 172PDB 93

Bowmag 75S 33 Dynapac LR100 42

Bowmag 90S 39 Dynapac 2100V 93

Bowmag 75AD 20 Dynapac CA121D 53

Bowmag 100AD 20 Dynapac CA121PD 54

Bowmag 120AD 34 Dynapac CA151 80

Bowmag 130AD 36 Dynapac CA151D 80

Bowmag BW122D 30 Dynapac CA151PD 96

EARTH PRESSURE DISTRIBUTION

TYPICAL VALUES GIVEN IN TABLE

SASKATCHEWAN AVENUE OVER STURGEON CREEKLATERAL EARTH PRESSURE INDUCED BY COMPACTION

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FIGURE 5

SEPTEMBER 2016

Amec Foster Wheeler

TETRA TECH WEI INC.DATE:

WX17718PROJECT No.:

PROJECT No.:

VA

LU

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F n

= z

/H

VALUE OF

0.2 0.4 0.6 0.8 1.0 0 0.5 1.0 1.5 2.001.0

0.8

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0.4

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m

0.1 0.60H

0.3 0.60H

0.5 0.56H

0.7 0.48H 0.48H0.6

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0.78

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=H 1.28m²n (m² + n² )²

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z = nH

x = mH

a

WA

LL

SECTION a-a

x = mH

PRESSURES FROM LINE LOAD(BOUSSINESQ EQUATION MODIFIED BY EXPERIMENT)

PRESSURES FROM POINT LOAD(BOUSSINESQ EQUATIONMODIFIED BY EXPERIMENT)

hP

RESULTANT Ph

Ph

QL

0.64 QL

Ph = 0.55 QL

FOR m > 0.4:

LINE LOAD QL FOR m ≤ 0.4:

QL

POINT LOAD QP

Ph

hP

QP

FOR m ≤ 0.4:

H²QP

H²QP

FOR m > 0.4:

0.28n² (0.16 + n²)³

1.77m²n² (m² + n² )³

QP

hPQP

hP Ph

QL

POINTLOAD

a

SASKATCHEWAN AVENUE OVER STURGEON CREEKLATERAL PRESSURE DUE TO

SURCHARGE POINT AND LINE LOADS

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FIGURE 6

SEPTEMBER 2016

Amec Foster Wheeler

TETRA TECH WEI INC.DATE:

WX17718PROJECT No.:

PROJECT No.:

20, 0

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25, 0

.62

0.0

0.1

0.2

0.3

0.4

0.5

0.6

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6



APPENDIX A



APPENDIX B



APPENDIX C

Photo 1: RW01 - Westbound lane on Saskatchewan Ave., ~1.0 m from curb

Photo 2: RW02 - Eastbound lane on Saskatchewan Ave., ~1.0 m from curb

C



C



APPENDIX D

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APPENDIX E

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031

2.53

396

602.

4266

485

3.15

448

1057

.222

4312

15.2

5468

1332

.986

9114

18.1

8932

1478

.564

4915

20.7

1097

1549

.825

7915

69.7

9304

1583

.418

8115

92.6

8555

1598

.973

2116

03.2

3281

1606

.115

439

11

68

79

810

46

57

5.4

7.4

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Load Intensity, p (kN/m)

Late

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· Project File No. WX17918 Amec Foster Wheeler Preliminary Design Geotechnical Report Environment...

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