Post on 07-Jul-2020
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Progressive Collapse Testing and Analysis of a Steel Frame Building
Halil Sezen1, Brian I. Song
2, and Kevin A. Giriunas
3
Abstract
A steel frame building was tested by physically removing four first story columns from
one of the perimeter frames prior to building’s scheduled demolition. The purpose of the field
experiment was to simulate sudden column loss in buildings that may cause progressive collapse.
Another objective was to investigate the load redistribution within the building after each column
removal. The measured experimental data and observed performance of the building was
valuable because it is very difficult and cost-prohibitive to build and test three-dimensional full-
scale building specimens in the laboratory. Generally, the design code requirements prescribe
simplified analysis procedures involving instantaneous removal of certain critical columns in a
building. Design methodologies and simplified analysis procedures recommended in the design
guidelines were also evaluated using the experimental data. In this study, two and three-
dimensional models of the building were developed and analyzed to simulate the progressive
collapse response. The effectiveness of the analysis procedures were evaluated by comparing
with the experimental data.
Introduction
Progressive collapse is a chain reaction of failures initiated by instantaneous loss of one
or a few vertical load carrying elements. Once the vertical structural element fails, the structure
should enable an alternative load-carrying path and transfer the loads carried by that element to
neighboring elements. Dynamic internal forces in adjoining members increase as a result of
release of internal energy due to a member loss. After the load is redistributed throughout the
structure, each structural component supports different loads including the additional internal
forces. If any redistributed load exceeds the capacities of surrounding undamaged members, it
1 Associate Professor, Department of Civil and Environmental Engineering and Geodetic Science, The Ohio State
University, Columbus, Ohio, e-mail: sezen.1@osu.edu (corresponding author) 2 Structural Engineer, P.E., M.S.C.E., URS Corporation, Warrenville, Illinois, e-mail: Brian.Song@urs.com
3 Graduate Student, Department of Civil and Environmental Engineering and Geodetic Science, The Ohio State
University, Columbus, Ohio, e-mail: giriunas.1@osu.edu
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can cause another local failure. Such sequential failures can spread from element to element,
eventually leading to the entire or a disproportionately large part of the structure. In general, such
progressive collapse happens in a matter of seconds.
The definition of progressive collapse may incorporate the concept of disproportionate
collapse, meaning that the extent of final failure is not proportional to the initial triggering events.
For example, the American Society of Civil Engineer (ASCE) Standard 7 defines progressive
collapse as "the spread of an initial local failure from element to element resulting eventually in
the collapse of an entire structure or a disproportionately large part of it" (ASCE 7, 2005). A
similar definition of progressive collapse is provided in the General Services Administration
(GSA) guidelines (2003): “a situation where local failure of a primary structural component
leads to the collapse of adjoining members, and hence, the total damage is disproportionate to the
original cause”.
The Ronan Point apartment tower collapse on May 16, 1968 is the first well-known case
of disproportionate progressive collapse (Griffiths et al., 1968). The building was a 22-story
precast concrete bearing wall system, located in Newham, England. The collapse was initiated by
a gas stove leak in a corner kitchen on the 18th floor. The Ronan Point collapse prompted
interest and concern in the structural engineering community all around the world. In particular,
this collapse led to significant changes in building codes in England and Canada to prevent
progressive collapse. Since the collapse of the World Trade Center (WTC) towers due to terrorist
attacks on September 11, 2001, interest in progressive collapse has further increased (NIST,
2005). The collapse of twin WTC structures was caused by a very large impact force and
subsequent fire. It was a progressive collapse and not a disproportionate collapse (Dusenberry et
al., 2004). This collapse showed that well-designed and robust modern buildings can also be
susceptible to progressive collapse.
Failure of one or more columns in a building and the resulting progressive collapse may
be a result of a variety of events with different loading rates, pressures or magnitudes. The
magnitude and probability of natural and man-made hazards are usually difficult to predict.
Therefore, current progressive collapse design guidelines are generally threat-independent and
do not intend to prevent such local damage. Rather, their purpose is to provide a level of
resistance against disproportionate collapse and to increase the overall structural integrity.
Design guidelines typically require minimum level of redundancy, strength, ductility and
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element continuity. In general, the design code requirements prescribe simplified analysis
procedures involving instantaneous removal of certain critical columns in a building. Among a
number of building codes, standards, and design guidelines for progressive collapse, General
Services Administration (GSA, 2003) and Department of Defense (DOD, 2005) address
progressive collapse mitigation explicitly. They provide quantifiable and enforceable procedures
to resist progressive collapse. This paper investigates the effectiveness of such commonly used
progressive collapse evaluation and design methodologies through numerical simulation and
experimental testing of a building.
A large number of analytical studies have been conducted to evaluate the effectiveness
and consistency of the current progressive collapse design guidelines. However, very limited
experimental research has been performed to validate the results of these computational studies
and to verify the methodologies prescribed in the guidelines. This is mainly because it is difficult
to construct and test full scale building specimens because such large scale testing is
discouragingly expensive.
In this study, several first-story columns were physically removed from an existing steel
frame building scheduled for demolition. The building was instrumented and experiment was
conducted prior to its demolition. Two and three-dimensional models of the building were
analyzed following the requirements of the current progressive collapse evaluation and design
guidelines, such as ASCE 7 (2005) and GSA (2003).
Progressive Collapse Design Approaches
Indirect and direct methods are the two approaches typically used for providing resistance
against progressive collapse (Ellingwood and Leyendecker, 1978). The indirect design approach
attempts to prevent progressive collapse through the provision of minimum levels of strength,
continuity, and ductility (ASCE 7, 2005). The examples of this approach are to improve
connection resistance by special detailing, to improve redundancy, and to provide more ductility
to a structure. The indirect design approach can easily be integrated into building codes or
standards because it can create a redundant structure that will perform well under various
extreme loading conditions and improve overall structural response. However, this method is
not recommended if the specific design goal is to prevent progressive collapse because of no
special consideration of specific loads or removal of critical vertical load carrying members. The
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goal of indirect design method or inclusion of general structural integrity requirements in design
codes or guidelines is to improve the overall structural performance of the building, not
specifically the progressive collapse resistance.
The direct design approach explicitly considers resistance of a building to progressive
collapse during the design process (ASCE 7, 2005). There are two direct design methods:
specific local resistance method and alternate load path method. The specific local resistance
method seeks to provide strength to be able to resist progressive collapse. The alternate load path
method seeks to provide alternative load paths to absorb localized damage and resist progressive
collapse.
The specific local resistance method requires that a critical structural element be able to
resist abnormal loading. Regardless of the magnitude of the load, the structural element should
remain intact because of its robustness. For this method, a sufficient strength and ductility must
be determined for the element during design against progressive collapse. Critical load carrying
elements can be designed to have additional strength and toughness to resist the abnormal
loading, simply by increasing the design load factors.
In the alternate path method, the design allows local failure to occur, but seeks to prevent
major collapse by providing alternate load paths. Failure in a structural member dramatically
changes load path by transferring loads to the members adjacent to the failed member. If the
adjacent members have sufficient capacity and ductility, the structural system develops alternate
load paths. Using this method, a building is analyzed for the potential of progressive collapse by
instantly removing one or several load-bearing elements from the building, and by evaluating the
capability of the remaining structure to prevent subsequent damage. The advantage of this
method is that it is independent of the initiating load, so that the solution may be valid for any
type of the hazard causing member loss. The alternate load path method is primarily
recommended in the current building design codes and standards in the U.S., including General
Services Administration (GSA, 2003) and the Department of Defense (DOD, 2005) guidelines.
Test Building and Experiment
The Bankers Life and Casualty Company (BLLC) insurance building was a three story
structure located in Northbrook, Illinois. The building was constructed in 1968 and was
demolished in August 2008 immediately after the experiment was conducted. The sixth edition
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of the AISC Steel Construction Manual (1963) was used to design the structure. The longitudinal
perimeter frame located on the north side of the BLCC building was tested and used in this paper
(Figure 1).
The tested part of the BLCC building was a two-story steel frame structure. As shown in
Figure 2, the heights of first and second floors were 20 ft-6 in. and 14 ft-8 in., respectively. The
building had a 10 ft-6 in. tall reinforced concrete framed basement. Although the experiment was
performed in the first story, the computer models included basement of the building, which may
have had very limited or no effect on the response of the upper two floors. Steel columns with
yield strength of 36,000 psi were rigidly connected to the concrete columns at the ground level.
Steel girders and beams had a specified yield strength of 42,000 psi. Test frame geometry and
designation of columns and beams of the BLCC building are shown in Table 1 and Figure 2.
A large loading dock bay area was located at the far Northwest corner of the structure. In
addition, there was a mezzanine level between the first and second floors at the Southeast and
Southwest end of the structure. For this research, the docking bay and mezzanine level were not
considered in computer modeling because neither would affect the experiment. Also, the plans
available do not include a recent addition of two stairwells at the south end of the building. This
addition does not have any effect on the experiment. Detailed description of the building, details
of instrumentation and testing, and building plans can be found in Giriunas (2009). Modeling
assumptions and details of structural analysis and results are provided in Song (2010) and Song
et al. (2010).
Four of the ten first-story columns were removed from the perimeter frame in the
following order: (1) two columns near the middle of the frame, (2) column in the building corner,
and (3) column next to the corner column (Figure 1). As shown in Figure 3a, a 3-ft long section
of the test columns was first torched at approximately 6 to 9 ft above the base of the column.
Only a very small portion of the flange was left intact when the two column cross sections were
cut. The column segment between the torched sections was then pulled out using a steel chain
(Figure 3b). No significant damage was observed in the building during or after the columns
were removed.
The demolition team first exposed the columns and beams by partially removing the
exterior brick walls. Strain gauges were attached on the columns numbered 5, 8, 11, 17 and 20 in
the frame model shown in Figure 4a. Strain gauges were typically installed on columns
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approximately 6 ft from ground. Universal general purpose strain gauges with a resistance of
120±0.3% Ohms were used. The major objective of strain gauge instrumentation was to monitor
the redistribution of loads during the removal of columns. A scanner and portable data
acquisition system were connected to a laptop computer to record the strains during testing.
Figure 5 shows the history of measured strain data for strain gauge 7, attached on column
11, during the torching and removal of columns. As shown in Figure 5, the strain values dropped
most after each torching or poking process. It is clear from the recorded data that there was a
sudden compressive (negative) strain increase of approximately 20 to 40x10-6
and 75 to 105x10-6
in column 11 near the end of torching of neighboring columns 14 and 17, respectively (see
Figure 4a). This indicates that part of the axial loads from these columns was transferred to
column 11. Details of the measured test data are reported by Giriunas (2009).
Modeling Assumptions and Structural Models
Several assumptions were made to simplify and clearly demonstrate the steps for
progressive collapse analysis of the two-dimensional (2-D) and three-dimensional (3-D) building
frame models. The assumptions are: 1) the building was modeled as a special moment resistant
frame with connections stronger than beams. Thus, the model allowed plastic hinges to form in
the beams, not in the connections or columns; 2) connections at the foundations were modeled as
pinned connections; 3) secondary members (e.g., transverse joist beams and braces) were
disregarded. Other than transferring the initial gravity loads, they did not directly contribute to
the progressive collapse resistance; 4) effect of large deflections was not considered. This is a
reasonable assumption in this study because very large deflections or collapse was not observed
in the test building; and 5) live load was assumed to be zero because non-structural loads were
removed from the building prior to its demolition.
At the time of testing, the frames carried only dead loads due to weight of walls, slabs,
beams and columns. The weight and properties of frame members were obtained from the
original design notes and structural drawings of the building. The weight of roof including
corrugated steel plates, membranes and roof joists was assumed to be 25 lb/ft2. The 12 in. thick
wall contained glass, brick, and concrete masonry units. To calculate the dead load of the walls,
the densities of glass, reinforced concrete masonry blocks, and exterior bricks were assumed to
be 160 lb/ft3, 135 lb/ft
3, and 120 lb/ft
3, respectively.
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The computer program SAP2000 (2010) was used to evaluate the progressive collapse
performance of the test building. Figure 4a shows 2-D model of the longitudinal perimeter frame
of the BLCC building. Four circled columns were sequentially removed in the SAP2000 analysis,
in the same order as the field torching and removal process. Figure 4b shows the 3-D model of
the building. The north side of the BLCC building, mainly considered in this study, had nine
bays in the longitudinal direction and eight bays in the transverse direction. To simplify 3-D
models, insignificant six bays in the back side were neglected. As shown in Figure 4b, the 3-D
model includes only front two bays that were most impacted by column removals.
Numerical Analysis and Results
Static or dynamic analysis methods with varying complexities, for example, including the
effect of geometric or material nonlinearities, can be used to analyze a structure. Researchers
investigated the advantages and shortcomings of different analysis procedures for progressive
collapse analysis (Marjanishvili 2004, Marjanishvili and Agnew 2006, Powell 2005). A complex
analysis is desired to obtain more realistic results representing the actual nonlinear and dynamic
collapse response of the structure. However, linear static analysis is presented in the GSA
guidelines (2003) as the primary method of analysis because it is cost-effective and easy to
perform. Nevertheless, it is difficult to predict accurate behavior in a structure due to the absence
of dynamic effects and material nonlinearity by sudden loss of one or more members
(Kaewkulchai and Williamson, 2003). The analysis is run under the assumptions that the
structure only undergoes small deformations and that the materials respond in a linear elastic
fashion. Linear static procedure, therefore, is limited to simple and low- to medium-rise
structures (i.e., less than ten stories) with predictable behavior (GSA, 2003).
In the linear static analysis performed in this research, dead loads were multiplied by 2.0
as recommended in the GSA guidelines (2003). The amplification factor of 2.0 is used to account
for dynamic effects, such as damping and inertia, when static analysis procedure is used. Results
of linear static analysis are evaluated here by comparing the demand-to-capacity ratios (DCR)
based on the recommendations of GSA guidelines. DCR for a structural component is defined as
the ratio of the maximum demand, e.g., moment, Mmax of the beam or column to its expected
capacity, e.g, ultimate moment capacity Mp, which is calculated as the product of plastic section
modulus and yield strength. In Mp calculations for columns, the effect of the axial load is
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neglected because the column axial loads were relatively small and did not significantly affect
the moment capacity of the cross section.
max
p
MDCR
M= (1)
If a DCR value is greater than 1.0, theoretically the member has exceeded its ultimate
capacity at that location. However, this alone does not signify failure of the structure as long as
other members are capable of carrying the forces redistributed after the initial plastic hinge
formation or failure. According to GSA (2003), if DCR values for steel columns and beams in the
BLLC building frame exceed 2.0 and 3.0, respectively, the members are to be considered failed
members, resulting in severe damage or potential partial or total collapse of the structure. The GSA
(2003) acceptance criteria is applied for DCR values calculated from static analysis, when only one
first floor corner column or only one first floor column at or near the middle of one of the perimeter
frames is removed from the computer model.
Table 2 and Figure 6 show DCR values for all members of the original 2-D frame and
after each of the four columns are removed. Frame member numbers up to 26 are columns, and
beams are numbered from 27 to 49 (Figure 4a). Figure 6 also shows the DCR limits for frame
members specified in the GSA (2003) for critical column removal scenarios. As shown in the
figure, calculated DCR values were quite large and exceeded the acceptance criteria. It should be
noted, however, that GSA criteria is applicable for removal of a single column. As indicated by
red inverted triangles in Figure 6, after the loss of the first column, only one member (column 18
with a DCR of 2.10) exceeded the GSA acceptance criteria.
After two or more columns were removed, many columns and beams, mostly above or
next to the first two removed columns, exceeded the DCR criteria. The DCR values for all
members remarkably increased after the last two columns were lost. The maximum calculated
DCR values for columns and beams were 6.50 and 8.03, respectively. When all four columns
were removed, the DCR values of almost all structural members were much higher than the
specified limits. Again, the GSA limits are specified for one column removal scenario. In
addition, in the computer model, the effect of infill walls was not considered. However, as shown
in Figure 1, the infill walls were mostly intact, probably carrying a significantly large portion of
the load and reducing the actual demand on the steel frame members. Also, DCR values were
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calculated from static analysis of the building frame with dead loads multiplied by 2.0, which
follows the GSA guidelines (2003).
DCR values reported in Table 2 and Figure 6 show that the beams were more impacted
than columns by the loss of columns. DCR values observed in beams were larger than in
columns for all column removal cases. This was probably due to large span lengths and large
slab tributary areas for the beams. The centerline distance between the transverse bays was 47 ft
in the first and second floors of the BLCC building. After a column is lost, the demand on the
beam bridging over the removed column significantly increases because the new beam spanning
over two bays has a much longer span length.
Figure 7 compares the DCR values for moments calculated from 2-D and 3-D models
after all four columns were removed. It was observed that DCR values calculated from the 3-D
model were smaller than those from 2-D model for almost all frame members. This simply could
be due to contribution of transverse beams, which enable the structure to have more stiffness, as
well as additional loads to be transferred to the columns in the transverse direction, leading to
decreased demands on the perimeter frame. Thus, it can be concluded that the 2-D model may
lead to overestimated demands. Figures 8 and 9 show the 2-D and 3-D models of the BLCC
building with corresponding DCR values, respectively, after the loss of four columns. Since no
significant visible damage was observed during the field testing, apparently the actual demands
were not as large as those predicted by the SAP2000 elastic static analysis.
Strains calculated from static analysis of the 2-D and 3-D models are compared with the
average strain measured by strain gauge 7 attached on Column 11 after the removal of each of
the four columns. Strains were calculated by considering the combination of axial loading and a
bending moment generated from 2-D and 3-D SAP2000 analyses. Details of strain calculations
from SAP2000 models were described in Giriunas (2009) and Song (2010). Figure 10 shows
comparison between calculated strains and strain data recorded in the field. 2-D SAP2000 model
significantly overestimated the measured response of the structure. The percent error was
calculated for each column removal case, indicating that the results obtained from 3-D model
were in close agreement with experimental results than 2-D model. Average 30% difference
between the 3-D model and the field data was observed while the average difference between the
2-D model and the field data was 84%.
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The loss of columns significantly affects the adjacent members, causing deformation of
the structure, especially in the area where columns are lost. Vertical displacements of the joints
right above the removed columns of the BLCC building were calculated. Figures 11 shows
changes in the joint displacements calculated from the 3-D analysis, during the entire column
removal process. Joints 1, 2, 3, and 4 designate the joints above the first, second, third, and
fourth removed columns, respectively (columns 14, 17, 5 and 2 in Figure 4a). Joints 1 and 2
above the first two removed columns had the largest displacements. This result was consistent
with the DCR values and 3-D deformed shape of the BLCC building shown in Figure 9.
Conclusions
Progressive collapse performance of an existing steel frame building was evaluated by
physically removing four first-story columns from the BLLC building and by analyzing the 2-D
and 3-D models of the building. The following conclusions were reached during this study based
on the evaluation of experimental data and structural analysis of the test building.
Most structural members of BLCC building exceeded the DCR limits once the second
column was removed. However, the building did not experience a collapse during the field test,
even after four columns were removed. Failure to consider the infill walls effect and the
amplification factor of 2 for the dead load in linear static analysis may have led to conservative
analysis results. Lower DCR values were observed in both columns and beams from the 3-D
model, compared with 2-D model. Force demand on each member of 3-D model is generally
smaller mainly because the additional loads caused by the loss of load-bearing columns were
distributed to more structural members in the 3-D model.
Strain data obtained from each field test was compared with the strain values calculated
from 2-D and 3-D models. The strain values from the 3-D model were closer to experimental
results. It was indicated that 3-D computer models were more accurate to simulate response of
buildings to removal of columns because 3-D models can account for 3-D effects including the
contribution of transverse members resulting in more conservative solutions.
Acknowledgements
This research was partially funded by the American Institute of Steel Construction. This
support and Research Director, Tom Schlafly’s feedback are gratefully acknowledged. The
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authors would like to thank the Environmental Cleansing Corporation of Markham, Illinois for
their help with the experiment.
References
AISC (1969), Manual of Steel Construction. 6th Edition, American Institute of Steel
Construction, Chicago, IL.
ASCE (2005), Minimum Design Loads for Buildings and Other Structures. American Society of
Civil Engineers (ASCE), Reston, VA.
DOD (2005), Design of Buildings to Resist Progressive Collapse. Unified Facilities Criteria
(UFC) 4-023-03, Department of Defense (DOD), Washington, D.C.
Dusenberry, D., Cagley, J. and Aquino, W. (2004), Case Studies. Multihazard Mitigation
Council National Workshop on Best Practices Guidelines for the Mitigation of Progressive
Collapse of buildings. National Institute of Building Sciences (NIBS). Washington, D.C.
Ellingwood, B. and Leyendecker, E. (1978), “Approaches for Design against Progressive
Collapse,” Journal of Structural Division, ASCE. Vol. 104, No. 3, pp. 413-423.
Giriunas, K. (2009), Progressive collapse analysis of an existing building. Undergraduate
Honors Thesis. Department of Civil and Environmental Engineering and Geodetic Science.
The Ohio State University, Columbus, OH.
Griffiths, H., Pugsley, A. and Saunders, O. (1968), Report of inquiry into the collapse of flats at
Ronan Point, Canning Town. Ministry of Housing and Local Government. London, United
Kingdom.
GSA (2003), Progressive Collapse Analysis and Design Guidelines for New Federal Office
Buildings and Major Modernization Projects. General Services Administration (GSA),
Washington, D.C.
Kaewkulchai, G. and Williamson, E.B. (2003), “Dynamic Behavior of Planar Frames during
Progressive Collapse,” 16th ASCE Engineering Mechanics Conference, July 16-18,
University of Washington, Seattle.
Marjanishvili, S.M. (2004), “Progressive Analysis Procedure for Progressive Collapse,” Journal
of Performance of Constructed Facilities, ASCE, Vol. 18, No. 2, pp. 79-85.
Marjanishvili, S. and Agnew, E. (2006), “Comparison of Various Procedures for Progressive
Collapse Analysis,” Journal of Performance of Constructed Facilities, ASCE. Vol. 20, No. 4,
pp. 365-374.
NIST (2005), The Collapse of the World Trade Center Towers, Final Report. National Institute
of Standards and Technology (NIST), Gaithersburg, MD.
Powell, G. (2005), “Progressive Collapse: Case Study Using Nonlinear Analysis.” ASCE
Structures Congress and Forensic Engineering Symposium, April 20-24, New York, NY.
SAP 2000 (2010), SAP 2000 Advanced Structural Analysis Program, Version 12. Computers and
Structures, Inc. (CSI). Berkeley, CA
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Song, B.I. (2010), “Experimental and Analytical Assessment on the Progressive Potential of
Existing Buildings.” Master’s Thesis. The Ohio State University. pp. 125
Song B., Sezen H., and Giriunas K. May 12-15, 2010. “Experimental and Analytical Assessment
on Progressive Collapse Potential of Actual Steel Frame Buildings.”ASCE Structures
Conference and North American Steel Construction Conference, American Society of Civil
Engineers, Orlanda, Florida
Table 1. Column and beam sections of the BLLC building
Column section Beam section
Column
number Column type
Beam
number Beam type
c1 concrete b1 RC flat slab
c2 10 WF 49 b2 24 I 79.9
c3 10 WF 72 b3 21 WF 62
c4 8 WF 31 b4 18 WF 45
Note: RC and WF are reinforced concrete and wide-flange shaped steel I-section, respectively. The first
and last numbers are the depth (inch) and nominal weight (lb per linear ft) of the steel column or beam,
respectively.
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Table 2. DCR values calculated from 2-D model for steel frame members.
Frame
member No.
Before
removal
1 column
removed
2 columns
removed
3 columns
removed
4 columns
removed
2 0.47 0.46 0.43 2.25 Removed
3 0.11 0.19 0.99 1.67 2.81
5 0.55 0.57 0.61 Removed Removed
6 0.06 0.16 1.13 1.43 4.02
8 0.55 0.54 0.37 1.67 1.81
9 0.03 0.17 2.00 3.86 2.35
11 0.55 1.61 3.13 3.24 3.54
12 0.04 1.54 3.26 3.33 5.02
14 0.55 Removed Removed Removed Removed
15 0.03 0.35 5.15 4.95 6.50
17 0.59 1.56 Removed Removed Removed
18 0.07 2.10 5.22 5.57 3.75
20 0.61 0.67 3.90 3.98 3.92
22 0.61 0.63 0.92 0.90 1.09
24 0.62 0.62 0.59 0.60 0.85
26 0.38 0.39 0.26 0.24 0.45
28 0.23 0.21 0.42 2.45 1.13
29 0.14 0.11 0.51 2.27 1.17
31 0.21 0.31 0.68 3.18 2.60
32 0.09 0.29 0.70 2.66 1.79
34 0.19 1.25 3.17 3.66 3.96
35 0.10 1.18 4.01 4.76 4.42
37 0.19 2.46 4.99 4.85 5.15
38 0.09 2.18 4.90 4.71 5.00
40 0.20 2.42 7.95 8.03 7.91
41 0.14 2.19 3.29 3.40 3.21
43 0.22 1.24 6.15 6.27 5.86
45 0.23 0.36 4.60 4.66 4.55
47 0.24 0.26 0.92 0.94 0.90
49 0.31 0.31 0.45 0.44 0.51
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Figure 1. Four circled columns were removed the BLCC building during the experiment.
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Figure 2. Longitudinal end frame elevation of BLLC building, including beam and columns
sections (see Table 1).
(a) (b) (c)
Figure 3. (a) Torching of column 16C, (b) column ready to be pulled out, and (c) removal of
column 16C.
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(a)
(b)
Figure 4. (a) two-dimensional SAP2000 model with frame member numbers, and (b) three-
dimensional SAP2000 model (circled columns are removed in the order shown).
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Figure 5. Strain data from strain gauge 7 during testing of the BLCC building.
Column 14
torching
Columns poking Column 17
torching
Column 5
torching
Columns removal
Column 2
torching
Column 5
torching
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Figure 6. Change in DCR values of each frame member in the 2-D model for all column removal
cases.
Figure 7. Comparison of DCR values determined from 2-D and 3-D models after four columns
removal.
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Figure 8.: Moment diagram of the 2-D model with corresponding DCR values after loss of four
columns.
Figure 9. Deflected 3-D model with corresponding DCR values after loss of four columns.
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Figure 10. Comparison of measured and calculated strain values.
Figure 11. Maximum joint displacements for the 3-D model after each column removal.