Volume 6 – Geotechnical Manual, Site Investigation and ...

Volume 6 – Geotechnical Manual, Site Investigation and

Engineering Survey

Jabatan Pengairan dan Saliran Malaysia Jalan Sultan Salahuddin 50626 KUALA LUMPUR

GOVERNMENT OF MALAYSIA DEPARTMENT OF IRRIGATION

AND DRAINAGE

DID MANUAL Volume 6

March 2009 i

Disclaimer

Every effort and care has been taken in selecting methods and recommendations that are appropriate to Malaysian conditions. Notwithstanding these efforts, no warranty or guarantee, express, implied or statutory is made as to the accuracy, reliability, suitability or results of the methods or recommendations. The use of this Manual requires professional interpretation and judgment. Appropriate design procedures and assessment must be applied, to suit the particular circumstances under consideration. The government shall have no liability or responsibility to the user or any other person or entity with respect to any liability, loss or damage caused or alleged to be caused, directly or indirectly, by the adoption and use of the methods and recommendations of this Manual, including but not limited to, any interruption of service, loss of business or anticipatory profits, or consequential damages resulting from the use of this Manual.

DID MANUAL Volume 6

ii March 2009

Foreword

The first edition of the Manual was published in 1960 and was actually based on the experiences and knowledge of DID engineers in planning, design, construction, operations and maintenance of large volume water management systems for irrigation, drainage, floods and river conservancy. The manual became invaluable references for both practising as well as officers newly posted to an unfamiliar engineering environment. Over these years the role and experience of the DID has expanded beyond an agriculture-based environment to cover urbanisation needs but the principle role of being the country’s leading expert in large volume water management remains. The challenges are also wider covering issues of environment and its sustainability. Recognising this, the Department decided that it is timely for the DID Manual be reviewed and updated. Continuing the spirit of our predecessors, this Manual is not only about the fundamentals of related engineering knowledge but also based on the concept of sharing experience and knowledge of practising engineers. This new version now includes the latest standards and practices, technologies, best engineering practices that are applicable and useful for the country. This Manual consists of eleven separate volumes covering Flood Management; River Management; Coastal Management; Hydrology and Water Resources; Irrigation and Agricultural Drainage; Geotechnical, Site Investigation and Engineering Survey; Engineering Modelling; Mechanical and Electrical Services; Dam Safety, Inspections and Monitoring; Contract Administration; and Construction Management. Within each Volume is a wide range of related topics including topics on future concerns that should put on record our care for the future generations. This DID Manual is developed through contributions from nearly 200 professionals from the Government as well as private sectors who are very experienced and experts in their respective fields. It has not been an easy exercise and the success in publishing this is the results of hard work and tenacity of all those involved. The Manual has been written to serve as a source of information and to provide guidance and reference pertaining to the latest information, knowledge and best practices for DID engineers and personnel. The Manual would enable new DID engineers and personnel to have a jump-start in carrying out their duties. This is one of the many initiatives undertaken by DID to improve its delivery system and to achieve the mission of the Department in providing an efficient and effective service. This Manual will also be useful reference for non-DID Engineers, other non-engineering professionals, Contractors, Consultants, the Academia, Developers and students involved and interested in water-related development and management. Just as it was before, this DID Manual is, in a way, a record of the history of engineering knowledge and development in the water and water resources engineering applications in Malaysia. There are just too many to name and congratulate individually, all those involved in preparing this Manual. Most of them are my fellow professionals and well-respected within the profession. I wish to record my sincere thanks and appreciation to all of them and I am confident that their contributions will be truly appreciated by the readers for many years to come.

Dato’ Ir. Hj. Ahmad Hussaini bin Sulaiman, Director General, Department of Irrigation and Drainage Malaysia

DID MANUAL Volume 6

March 2009 iii

Table of Contents

Disclaimer .................................................................................................................................. i

Foreword .................................................................................................................................. ii

Table of Contents ...................................................................................................................... iii

List of Volumes ........................................................................................................................ iv

Part 1 GEOTECHNICAL MANUAL

Part 2 SITE INVESTIGATION

Part 3 ENGINEERING SURVEY

DID MANUAL Volume 6

iv March 2009

List of Volumes

Volume 1 FLOOD MANAGEMENT

Volume 2 RIVER MANAGEMENT

Volume 3 COASTAL MANAGEMENT

Volume 4 HYDROLOGY AND WATER RESOURCES

Volume 5 IRRIGATION AND AGRICULTURAL DRAINAGE

Volume 6 GEOTECHNICAL MANUAL, SITE INVESTIGATION AND ENGINEERING SURVEY

Volume 7 ENGINEERING MODELLING

Volume 8 MECHANICAL AND ELECTRICAL SERVICES

Volume 9 DAM SAFETY

Volume 10 CONTRACT ADMINISTRATION

Volume 11 CONSTRUCTION MANAGEMENT

DID MANUAL Volume 6

March 2009 i

Acknowledgements

Steering Committee: Dato’ Ir. Hj. Ahmad Husaini bin Sulaiman, Dato’ Nordin bin Hamdan, Dato’ Ir. K. J. Abraham, Dato’ Ong Siew Heng, Dato’ Ir. Lim Chow Hock, Ir. Lee Loke Chong, Tuan Hj. Abu Bakar bin Mohd Yusof, Ir. Zainor Rahim bin Ibrahim, En.Leong Tak Meng, En. Ziauddin bin Abdul Latiff, Pn. Hjh. Wardiah bte Abd. Muttalib, En. Wahid Anuar bin Ahmad, Tn. Hj. Zulkefli bin Hassan, Ir. Dr. Hj. Mohd. Nor bin Hj. Mohd. Desa, En. Low Koon Seng, En.Wan Marhafidz Shah bin Wan Mohd. Omar, Ir. Md Fauzi bin Md Rejab, En. Khairuddin bin Mat Yunus, Cik Khairiah bt Ahmad, Coordination Committee: Dato’. Nordin bin Hamdan, Dato’ Ir. Hj. Ahmad Fuad bin Embi, Dato’ Ong Siew Heng, Ir. Lee Loke Chong, Tuan Hj. Abu Bakar bin Mohd Yusof, Ir. Zainor Rahim bin Ibrahim, Ir. Cho Weng Keong, En. Leong Tak Meng, Dr. Mohamed Roseli Zainal Abidin, En. Zainal Akamar bin Harun, Pn. Norazia Ibrahim, Ir. Mohd. Zaki, En. Sazali Osman, Pn. Rosnelawati Hj. Ismail, En. Ng Kim Hoy, Ir. Lim See Tian, Ir. Mohd. Fauzi bin Rejab, Ir. Hj. Daud Mohd Lep, Tn. Hj. Muhamad Khosim Ikhsan, En. Roslan Ahmad, En. Tan Teow Soon, Tn. Hj. Ahmad Darus, En. Adnan Othman, Ir. Hapida Ghazali, En. Sukemi Hj. Sidek, Pn. Hjh. Fadzilah Abdul Samad, Pn. Hjh. Salmah Mohd. Som, Ir. Sahak Che Abdullah, Pn. Sofiah Mat, En. Mohd. Shafawi Alwi, En. Ooi Soon Lee, En. Muhammad Khairudin Khalil, Tn. Hj. Azmi Md Jafri, Ir. Nor Hisham Ghazali, En. Gunasegaran M., En. Rajaselvam G., Cik Nur Hareza Redzuan, Ir. Chia Chong Wing, Pn Norlida Mohd. Dom, Ir. Lee Bea Leang, Dr. Hj. Md. Nasir Md. Noh, Pn Paridah Anum Tahir, Pn. Nurazlina Mohd Zaid, PWM Associates Sdn. Bhd., Institut Penyelidikan Hidraulik Kebangsaan Malaysia (NAHRIM), RPM Engineers Sdn. Bhd., J.U.B.M. Sdn. Bhd. Working Group: Pn. Rozaini binti Abdullah, En. Azren Khalil, Tn. Hj Fauzi Abdullah, En. Che Mohd Dahan Che Jusof, En. Ng Kim Hoy, En. Dzulkifli bin Abu Bakar, Pn. Che Shamsiah bt Omar, En. Mohd Latif Bin Zainal, En. Mohd Jais Thambi Hussein, En. Osman Mamat, En. Tajudin Sulaiman, Pn. Rosilawani binti Sulong, En. Ahmad Solihin Budarto, En. Noor Azlan bin Awaludin, Pn. Mazwina bt Meor Hamid, En. Muhamad Fariz bin Ismail, Cik Sazliana bt Abu Omar, Cik Saliza Binti Mohd Said, En. Jaffri Bahan, En. Mohd Idrus Amir, Mej (R) Yap Ing Fun, Ir Mohd Adnan Mohd Nor, Ir Liam We Lin, Ir. Steven Chong, En. Jamal Abdullah, En. Ahmad Ashrin Abdul Jalil, Cik Wan Yusnira Wan Jusoh @ Wan Yusof.

DID MANUAL Volume 6

ii March 2009

Registration of Amendments

Amend No

Page No

Date of Amendment Amend

No Page No

Date of Admendment

DID MANUAL Volume 6

March 2009 iii

Table of Contents

Acknowledgements ..................................................................................................................... i

Registration of Amendments ...................................................................................................... ii


List of Symbols ......................................................................................................................... iv

Chapter 1 GENERAL

Chapter 2 GEOTECHNICAL DESIGN PROCESS

Chapter 3 FUNDAMENTAL PRINCIPLES

Chapter 4 SOIL SETTLEMENT

Chapter 5 BEARING CAPACITY THEORY

Chapter 6 SLOPE STABILITY

Chapter 7 RETAINING WALL

Chapter 8 GROUND IMPROVEMENT

Chapter 9 FOUNDATION ENGINEERING

Chapter 10 SEEPAGE

DID MANUAL Volume 6

iv March 2009

List of Symbols

Unit weight γ

d Dry unit weight γ

w Unit weight of water γ

γb Buoyant unit weight

S Degree of saturation

w Moisture content

e Void ratio

e0 Initial void ratio

n Porosity

G

σ Total stress

s Specific gravity of solids

u

σ’ Effective stress

Pore water pressure

g

ρw Density of water

Gravity

c Cohesion

Cc Compression Index

Cr Recompression Index

U Degree of consolidation

t

Angular distortion

Time

θ

δ Differential settlement in the structure

qult Ultimate net bearing capacity

qu

Frictional angle

Allowable net bearing capacity

’ Effective frictional angle

Ka Coefficient of active earth pressure

Kp Coefficient of passive earth pressure

Es Young’s modulus of soil

PART 1: GEOTECHNICAL MANUAL

CHAPTER 1 GENERAL

Chapter 1 GENERAL

March 2009 1-i

Table of Contents

Table of Contents ......................................................................................................... 1-i

1.1 PURPOSE AND SCOPE ....................................................................................... 1-1

1.2 LIMITATION OF MANUAL ................................................................................... 1-1

Chapter 1 GENERAL

1-ii March 2009

(This page is intentionally left blank)

Chapter 1 GENERAL

March 2009 1-1

1 GENERAL

1.1 PURPOSE AND SCOPE Part 1 Volume 6 is developed around the aspects of geotechnical engineering usually required in JPS nature of work, that include earth retaining structures, river works, embankment, revetment, slope stability and stabilization works as well as the various coastal and hydraulic related works. It serves to provide a very selective and by no means comprehensive overview of fundamental practical knowledge ranging from methods of theoretically based analysis to “rules of thumb” solutions for geotechnical and foundation analysis, design and construction issues encountered in JPS work. It is envisaged that this manual will most likely be used by practicing civil generalists, geotechnical and foundation specialists, and others involved in the planning, design and construction of JPS’s nature of works. The main goals of this Part are to:- a) Provide a general understanding and appreciation of the geotechnical principles gearing

towards a sound, safe and cost-effective design and construction of JPS projects. b) Serve as a consistent guidance for the practitioners involved in the geotechnical planning,

design and construction in all phases of a JPS project. c) Encourage the readers to follow through the topic of interest in one or more of the

reference books mentioned in the references 1.2 LIMITATION OF MANUAL Even though the material presented is theoretically correct and represents the current state-of-the-practice, the user must realize that there is no possible way to cover all the various intricate aspects of geotechnical engineering. Owing to the high degree of ambiguities and uncertainties in the various aspect of geotechnical engineering, sound engineering judgment from highly experience and competent specialist practicing engineer is most important. For example, the values for the parameters to be used in the analysis and design should be selected by a geotechnical specialist who is intimately familiar with the type of soil in that region and intimately knowledgeable about the regional construction procedures that are required for the proper installation of such foundations in local soils. Often the key in the successful practice and application of geotechnical engineering lies in a sound knowledge and understanding of the engineering properties and behavior of soils in situ when subjected to changes in the environment conditions such as engineering loading or unloading.

Chapter 1 GENERAL

1-2 March 2009


CHAPTER 2 GEOTECHNICAL DESIGN PROCESS


March 2009 2-i

Table of Contents

Table of Contents .................................................................................................................. 2-i

List of Tables ....................................................................................................................... 2-ii

List of Figures ...................................................................................................................... 2-ii

2.1 GENERAL ................................................................................................................. 2-1

2.2 DESIGN PROCESS ..................................................................................................... 2-1

2.2.1 Determine Type of Geotechnical Design and Parameters Required ................. 2-2

2.2.2 Decide on Appropriate Geotechnical Investigation ......................................... 2-5

2.2.3 Interpret Geotechnical Investigation Result to Obtain Representative Parameters/Properties ................................................................................ 2-5

2.2.4 Designer’s Analysis and Design ................................................................... 2-6

2.2.5 Check Compliance and Need for Modification during Construction .................. 2-6

2.2.6 Post Construction Monitoring and Verification of Structure Performance .......... 2-7

REFERENCES ....................................................................................................................... 2-8


2-ii March 2009

List of Tables

Table Description Page

2.1 Typical Scope of DID Works (After Geotechnical Guidelines for DID Works) 2-3

2.2 Type Of Geotechnical Analysis Corresponding To Design Component 2-3

List of Figures

Figure Description Page

2.1 Flow Chart for the Designer Involvement in Geotechnical Design 2-2

2.2 Some Typical DID's Structures 2-4

2.3 Combination of Sources of Information in Geotechnical Design 2-6


March 2009 2-1

2 GEOTECHNICAL DESIGN PROCESS

2.1 GENERAL Geotechnical engineering is highly empirical and is perhaps much more of an ‘art’ than the other disciplines within civil engineering because of the basic nature of soil and rock materials. They are often highly variable, heterogeneous and anisotropic i.e. their engineering and material properties may vary widely within the soil mass and also may not be the same in all direction. Furthermore, the behavior of soil and rock materials are often controlled by the joints, fractures, weak layers and zones and other ‘defects’ in the materials. In the application of geotechnical engineering, the soil is usually assumed to be homogenous and isotropic obeying linear stress-strain laws. However, to account for the real material behavior, large empirical correction or ‘factors of safety’ must be applied in geotechnical design. As such, geotechnical engineering is really an ‘art’ rather than an engineering science, where good judgment and practical experience of the designer and contractors are essential for a successful geotechnical design. 2.2 DESIGN PROCESS In geotechnical engineering, the analysis and design process normally involved the various steps as illustrated in Figure 2.1. It includes determination of the type of geotechnical design and their required parameters, identification of appropriate geotechnical investigation works, evaluation and interpretation of geotechnical investigation result to obtain representative parameters and properties, performing design and analysis, checking compliance during construction and post construction monitoring.


2-2 March 2009

DESIGNER ASSIGNED PROJECT

DETERMINE TYPE OF GEOTECHNICAL DESIGN AND PARAMETERS REQUIRED

DECIDE ON APPROPRIATE GEOTECHNICAL INVESTIGATIONS

INTERPRET GEOTECHNICAL INVESTIGATION RESULT TO OBTAIN REPRESENTATIVE PARAMETERS/PROPERTIES

DESIGNER’S ANALYSIS AND DESIGN

CHECK COMPLIANCE AND NEED FOR MODIFICATION DURING CONSTRUCTION

POST CONSTRUCTION MONITORING AND VERIFICATION OF STRUCTURE PERFORMANCE

Figure 2.1 Flow Chart for the Designer Involvement in Geotechnical Design

2.2.1 Determine Type of Geotechnical Design and Parameters Required The type of geotechnical analysis and design depends very much on the type of structures or works to be designed. Table 2.1 below highlighted the types of works normally carried out by DID and their associated design components which include various hydraulic structures; embankments and dams; subsurface drainage; excavations; earth retaining structures and revetment works. The type of geotechnical analysis required and corresponding to the design components are as in Table 2.2, namely bearing capacity, settlement, slope stability, seepage, retaining wall, soil and geosynthetic filter.


March 2009 2-3

Table 2.1 Typical Scope of DID Works (After Geotechnical Guidelines for DID Works)

Design

Components Scope of Work

Hydraulic Structure

Embankments and Dams

Sub-surface Drainage

Excavation Works

Retaining Structures Revetment

1. River Works and Erosion control

X X X

2. Irrigation and Drainage X X X X

3. Flood Mitigation X X X X

4. Urban Drainage X X X X X X

5. Coastal Engineering X X X

Table 2.2 Type Of Geotechnical Analysis Corresponding To Design Component

Geotechnical

Analyses Design Components

Bearing Capacity Settlement Slope

Stability Seepage Retaining wall

Soil and Geosynthetic

Filter

1. Hydraulic Structure X X X X X

2. Embankments and Dams X X X X

3. Retaining Structure X X X X X

4. Subsurface Drainage X X X

5. Excavations X

6. Revetments X X X

Some typical DID structures are as shown in Figure 2.2


2-4 March 2009

Figure 2.2 Some Typical DID's Structures


March 2009 2-5

2.2.2 Decide on Appropriate Geotechnical Investigation The objectives and various general details on the type of geotechnical investigation works are described in Part 2, Volume 6 : Soil Investigation which include both field and laboratory works. Suffice here to mention that the composition and amount of geotechnical investigation proposed shall be able to provide sufficient data on the ground, groundwater conditions at the proposed site and proper description of the essential soil properties for geotechnical design and construction. It shall also be planned to take into account the construction and performance requirements of the proposed structure. Very often geotechnical engineer is required to determine the type of soil investigation works in relation to the envisage analysis required in the design works, i.e. the long-term (drained with effective stress analysis) or short-term analysis (undrained total stress analysis) conditions. 2.2.3 Interpret Geotechnical Investigation Result to Obtain Representative Parameters/Properties The evaluation and interpretation of geotechnical investigation work shall include a review of the field and laboratory results to derive at the reasonable and representative parameters and properties. This normally involves tabulation and graphical presentation of field and laboratory results such as the range and distribution of values of the required soil parameters (including ground water condition), subsurface strata profile which differentiate and group the various formations and properties. Any irregularities or adverse field and laboratory results shall be pointed out, commented upon, and if necessary to propose further geotechnical investigation for verification. Reader should refer to Part 2 Volume 6 for more detail and comprehensive information on this topic. In spite of the many advances in geotechnical engineering theory, there are still many uncertainties in the analysis and design due mainly to the highly variable, heterogeneous and anisotropic nature of soil material. Designer normally use various investigation and testing techniques to determine the soil conditions, however even the most thorough investigation program encounters only a small portion of the soils and relies heavily on the interpolation and extrapolation. The most practical approach to solve geotechnical design issues is to combine the sources of information gathered through soil investigation and testing program, established theory developed to predict the behavior of soils and experience obtained from previous projects coupled with sound engineering judgment. These approaches are depicted in Figure 2.3


2-6 March 2009

Site Investigation/ laboratory Testing

Established Theory

Experience and Judgment

Figure 2.3 Combination of Sources of Information in Geotechnical Design 2.2.4 Designer’s Analysis and Design Some the common geotechnical analysis and design carried out by the Department include evaluation and determination of the soil bearing capacity, settlement, seepage forces; and stability of slope, earth retaining structures as well as the selection of effective soil and geosynthetic filter in sub-soil drainage.

In carrying out the analysis and design, sound engineering by experience geotechnical engineer should be incorporated to compensate for the many uncertainties in actual soil behavior, which should take into consideration the following factors: • Required reliability or acceptable probability of failure • Consequence of failure • Degree of uncertainties in soil properties and applied loads • Compromise between cost and reliability • Degree of ignorance of the structure behaviour 2.2.5 Check Compliance and Need for Modification during Construction During construction, site operation shall be checked for compliance with the method of construction assumed in the design. Also, observation and measurements of the structure and its surrounding may necessitate some remedial measures or alterations to the construction sequence, for example the unexpected excessive settlement of the embankment under construction would warrant the review of the design and proposed sequence of construction. In fact, a great deal of geotechnical information can be gathered during construction phase of a project, particularly those involving huge volume of earth excavation or exposure where the actual ground conditions can be identified. These information should then be used to validate the geotechnical design assumptions or soil parameters and if necessary, to revise and modify the design accordingly.


March 2009 2-7

2.2.6 Post Construction Monitoring and Verification of Structure Performance A geotechnical design should not be considered completed upon the completion of the construction works. The designer should also be involved in post-construction activities such as visual observation and inspection of the structure; gathering and analyzing results of instrumentation monitoring to ensure its long-term performance and identified any necessary maintenance work. Any lesson learned from the design stage to the completion of the construction works should be adequately documented for future references.


2-8 March 2009

REFERENCES [1] Bowles, J.E. Foundation Analysis and Design. (Fourth edition). McGraw-Hill International, New York, 1992, 1004 p. [2] Brown, R.W., (1996) Practical foundation Engineering Handbooks, Mcgraw-Hill [3] BSI. Eurocode 7: Geotechnical Design – Part 1: General Rules (BS EN 1997-1 : 2004). British Standards Institution, London, 2004, 117 p. [4] Carter M. & Symons, M.V., Site Investigations and foundations Explained, Pentech Press, London [5] CGS, Canadian Foundation Engineering Manual, (Third edition). Canadian Geotechnical Society, Ottawa, 1992, 512 p. [6] Das, B.M., Principles of Geotechnical Engineering, PWK-Kent Publishing Company , Boston,MA., 1990 [7] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C., NAVFAC DM-7.1, May 1982, Soil Mechanics [8] DID, Geotechnical Guidelines for D.I.D Works [9] Holtz, R.D., Kovacs, W.D. An Introduction to Geotechnical Engineering, Prentice-Hall, Inc. New Jersey [10] Koerner R.M .• Construction and Geotechnical Method in Foundation Engineering, McGraw Hill, 1985. [11] Lambe T.W. and Whitman R.V., Soil Mechanics, John Wiley 8: Sons, 1969 [12] Peck R.B Hanson W.E. and Thornburn R.H., “Foundation Engineering", John Wiley and Sons, 1974. [13] Smith C.N., Soil Mechanics for Civil and Mining Engineers. [14] Teng W.C., Foundation Design, Prentice Hall, 1984. [15] Terzaghi, K. & Peck, R.B. (1967). Soil Mechanics in Engineering Practice. (Second edition). Wiley, New York, 729 p.

CHAPTER 3 FUNDAMENTAL PRINCIPLES


March 2009 3-i

Table of Contents

Table of Contents .................................................................................................................... 3-i

List of Tables ......................................................................................................................... 3-ii

List of Figures ........................................................................................................................ 3-ii

3 FUNDAMENTAL PRINCIPLES ................................................................................................. 3-1

3.1 BASIC WEIGHT-VOLUME RELATIONSHIPS ..................................................................... 3-1

3.2 EFFECTIVE STRESS CONCEPT ....................................................................................... 3-2

3.3 VERTICAL STRESS DISTRIBUTION ................................................................................ 3-4

3.4 SHEAR STRENGTH ....................................................................................................... 3-5

3.4.1 Basic Principle................................................................................................. 3-5

3.4.2 Effective Versus Total Stress Analysis ............................................................... 3-8

REFERENCES ....................................................................................................................... 3-11


3-ii March 2009

List of Tables


3.1 Definition and Typical Values of Common Soil Weight-Volume Parameters 3-1

3.2 Some Unit Weight Volume Inter-Relationships 3-2

3.3 Design Conditions and Related Shear Strengths and Pore Pressures 3-10

List of Figures


3.1 Unit Soil Mass and Phase Diagram 3-1

3.2 Total Stress at a Point 3-2

3.3 Example 3.1 3-3

3.4 Schematic of the Vertical Stress Distribution with Depth under an Embankment generated by FoSSA Program (from Soil and Foundation - FHWA) 3-4

3.6 Graphical Representative of Shear Strength 3-7

3.7 Mohr-Coulomb’s Circles and Failure Envelopes 3-8


March 2009 3-1

3 FUNDAMENTAL PRINCIPLES

3.1 BASIC WEIGHT-VOLUME RELATIONSHIPS Soil mass is generally idealized as a three phase system consisting of solid particles, water and air as illustrated in diagram in Figure 3.1. Owing to the three different components of soils, complex states of stresses and strains may exist in a soil mass. The various volume changes phenomena encountered in geotechnical engineering, such as deformation, consolidation, collapse, compaction, expansion, shrinkage etc. can be described in term of the various volumes of these components in the soil mass. Thus, knowledge of the relative proportion of each component and their various inter-relationships can give an important insight into engineering behavior of a particular soil. The weight-volume relationships of the soil mass are readily available in most soil mechanics textbooks. Most of these relationships are as summarized in Table 3.1 and Table 3.2.

Soil particles Voids (filled with water and air)

Figure 3.1 Unit Soil Mass and Phase Diagram

Table 3.1 Definition and Typical Values of Common Soil Weight-Volume Parameters

Typical Range Parameter Symbol Definition English SI

Unit weight WV

90 – 130 lb/ft3 14 – 20 kN/m3

Dry unit weight d Ws

V 60 – 125 lb/ft3 9 – 19 kN/m3

Unit weight of water w Ww

V 62.4 lb/ft3 9.8 kN/m3

Buoyant unit weight b sat - w 28 – 68 lb/ft3 4 – 10 kN/m3

Degree of saturation S Vw

Vv x 100% 2 – 100% 2 – 100%

Moisture content w Ww

Ws x 100% 3 – 70% 3 – 70%

Void ratio e Vv

Vs 0.1 – 1.5 0.1 – 1.5

Porosity n Vv

V x 100% 9 – 60% 9 – 60%

Specific gravity of solids Gs Ws

Vs w 2.6 – 2.8 2.6 – 2.8

(Source: Donald P. Coduto, [6])

Air

Water

Solid

Volume Weight

VvVw

Va

Vs

Ww

Ws

Wa≈0

V

1 unit

W


3-2 March 2009

Table 3.2 Some Unit Weight Volume Inter-Relationships

Unit-weight Relationship Dry Unit Weight (No Water) Saturated Unit Weight (No Air)

t=1+w Gs w

1+e

t=Gs+Se w

1+e

t=1+w Gs w

+ wGs1 S

t=Gs w 1-n (1+w)

d=t

1+w

d=Gs t

+e1

=Gs w -n) d (1

t=Gs w

1+ wGs

S

=eS w

+d 1 e w

t‐n w d sa

d= sat-e

1+e

w

=Gs+e w

1 esat +

sat 1-n Gs+n w

sat=1+w

1+wGsGs w

satew

1+w1+e w

n sat d w

sat de

1+e

w

In above relations, w refers to the unit weight of water, 62.4 pcf (=9.81 kN/m3). (Source: Donald P. Coduto, [6]) 3.2 EFFECTIVE STRESS CONCEPT The concept of effective stress was first proposed by Karl Terzaghi in the mid sixties. It is a simple concept with significant implications on how the science of geotechnical engineering develops. In simple terms the concept stipulates that soil consists of 2 major components in general, i.e., (i) particulate, and (ii) pore water.

Under an applied load, the total stress (σ) in a saturation unit soil mass is composed of inter-granular stress and the pore water pressure (u) as illustrated in Fig 3.2. When pore water drains from the soil, the contact between the soil grains will increase which increases the inter-granular stress. The inter-granular stress is called the effective stress, σ’.

Figure 3.2 Total Stress at a Point

The concept of effective stress is extremely useful in the development of soil strength theories and soil behaviour models. It allows a better understanding of soil behaviour, interpreting laboratory test results and making engineering design calculations such as in the estimation of settlement due to consolidation. More significantly, the concept implies that the soil shearing strength depends only on the effective stress componentpore water carries no shear under hydrostatic or steady state seepage conditions (i.e., flow velocity is negligible).

Particles Pore WaterParticles Pore WaterParticles Pore Water Mathematically, σ = σ’ + u Where σ = Total stress σ' = effective stress u = pore water pressure


March 2009 3-3

Both the total stress and pore water pressure may readily be estimated or calculated with knowledge of the densities and thickness of soil layers and location of ground water stable. To calculate the total vertical stress σv at a point in a soil mass, you simply sum up the weights of all the material (soil solids + water) above that point multiplied by respective thickness of each soil layer or σv = ∑ ρi

ni= gzi (3.1)

σv = Vertical stress ρi = Densities of each layer above point in question g = Gravity z = Thickness of each layer n = Number of layers above point in question

The pore water pressure is similarly calculated for static water conditions i.e. u = ρw g zw (3.2)

Where ρw = density of water zw = depth below ground water table to the point in question Example: 3.1 Given that the container of soil shown in Fig 3.3 with the saturated density as 2.0 Mg/m3 Calculate the total and effective stress at Elevation A

Water

The stresses at Elevation A due to the submerged soil and water above are: Total stress = ρsat g h + ρw g z w = (2 x 9.81 x 5.0) + (1 x 9.81 x 2.0) = 117.7 kPa Pore water pressure, u = ρw g (z w + h) = 1 x 9.81 x (2 + 5) = 68.7 kPa Effective stress at Elev. A, σ ’ = σ − u = ( ρsat g h + ρw g z w ) - ρw g (z w + h) = 117.7 - 68.7 = 49.0 kPa

Soil

Zw = 2 m

h = 5 m

Elev. A

Figure 3.3 Example 3.1


3-4 March 2009

3.3 VERTICAL STRESS DISTRIBUTION When a very large area is to be loaded, the induced stress in underneath soil would be would be 100% of the applied stress at the contact surface. However, near the edge or end of the loaded area you might expect a certain amount of attenuation of stress with depth because no stress is applied beyond the edge. Likewise, with a footing of limited size the applied stress would dissipate rather rapidly with depth. Figure 3.4 illustrated a schematic of the vertical stress distribution with depth along the center line under an embankment of height, h, constructed with a soil having total unit weight, γ t .

Figure 3.4 Schematic of the Vertical Stress Distribution with Depth under an Embankment generated by FoSSA Program (from Soil and Foundation - FHWA)

One of the simplest methods to compute the distribution of stress with depth for a loaded area is to use the 2 to 1 (2:1) method. This is an empirical approach based on the assumption that the area over which the load acts increases in a systematic way with depth. Since the same vertical force is spread over an increasingly larger area, the unit stress decreases with depth, as shown in Fig. 3.4. In Fig. 3.5a, a strip or continuous footing is seen in elevation view. At a depth z, the enlarged area of the footing increases by z/2 on each side. The width at depth z is then B + Z and the stress σz at that depth is

σz = load

B+z ×1 =

σo(B×1)(B+z)×1

(3.3)

By analogy, the corresponding stress at depth z for a rectangular footing of width B and length L (as illustrated in Figure 3.5b would be

∆σz = load

B+z (L+z) =

σoBLB+z (L+z)

(3.4)

March 2

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3-6 March 2009

The shear strength of soil can be may be expressed by Coulomb’s equation: s = c + σ tan φ (3.5) where s = shear strength or shear resistance c = cohesion φ = angle of internal friction of soil σ = total normal stress to shear plane For effective stresses the shear strength is expresses as: s = c '+ σ' tan φ' and (3.6) σ' = (σ − u) (3.7) where c' = effective cohesion φ' = effective angle of internal friction σ' = effective stress or inter-granular stress normal to the shear plane u = pore water pressure on the shear plane The equation 3.1 and 3.2 could also be represented graphically in Figure 3.6. As expressed in the above equations, the shear strength of soil is represented by the additive of two terms i.e. σ tan φ (οr σ'tan φ) and c (or c’). The first term is the inter-granular frictional component which is approximately proportional to the normal stress on the surface, σ (or σ'), whereas the second term is due to the internal electro-chemical bonding between particles and is independent of the normal stress. A coarse-grained soil such as sand and gravels has no cohesion and thus, it strength depends solely on the inter-granular friction between soil grains. This type of soil is called granular, cohesionless, non-cohesive or frictional soil. On the other hand, soils containing large amounts of fine grains (clay, silt and colloid) are called fine-grained or cohesive soils.


March 2009 3-7

Figure 3.6 Graphical Representative of Shear Strength The shear strength parameters, c and σ or c' and σ ', are normally determined from laboratory shear test results such as triaxial and direct shear tests. A series of tests are usually carried out whereby the stresses (normal and shear stresses) from each test representing failure are plotted. The resulting graph, as illustrated in Figure 3.7, is known as the Mohr-Coulomb (M-C) failure envelope which represents the shear strength of the soil.


3-8 March 2009

Figure 3.7 Mohr-Coulomb’s Circles and Failure Envelopes The physical meaning of the M-C failure envelope may be explained as follows: • Every point on the M-C failure envelope represents a combination of normal and shear stress

that results in failure of the soil, i.e. the limiting state of stress for equilibrium. • If the state of stress is represented by a point below the M-C failure envelope then the soil

will be stable for that state of stress. • States of stress beyond the M-C failure envelope cannot exist since failure would have

occurred before that point could be reached. 3.4.2 Effective Versus Total Stress Analysis It is important to note that the properties of soil and its shear strength in the vicinity of construction facility could change with time. As explained in Item 3.2, when the stress in the soil is suddenly changed (e.g. due to applied load), the additional stress is initially carried by the pore water pressure resulting to what is known as excess pore water pressure. If a foundation consolidates slowly, relative to the rate of construction, a substantial portion of the applied load will be carried by the pore water, which has no shear strength, and the available shearing resistance is limited to the in-situ shear strength. In this case, analysis are carried out using the total stress (undrained) analysis.

M-C Failure Envelope

M-C Failure Envelope


March 2009 3-9

In time , the excess pore water pressure will dissipate as result of seepage under consolidation and the stress is eventually carried by soil skeleton of the soil and under such condition, analysis using the effective (drained) stress analysis is applied. Since shear strength will vary with time, it is important for the designer to understand and determine at which point in time i.e. before, during or after construction that is critical to the design of the structure. As granular or sandy soils are more permeable than cohesive or clayey soils, drainage of excess pore pressure in sandy soil occurs much more rapidly. Hence, effective (drained) stress analysis is usually necessary for sandy soils. For clayey soil, either a total (undrained) stress analysis or effective (drain) stress analysis is required depending on the time considered in relation to the duration of construction. Effective stress analysis requires the estimation of the drained strength parameters c’, φ’ and pore pressures. However, with pure free draining sands, φ = φ’ and c = 0. For total stress analysis, undrained parameters typically used are φ = 0 and c determined from in-situ vane shear (for soft clay) or undrained unconfined (UU) and consolidated undrained (CIU) triaxials tests. In general, depending on the soil compressibility, thickness, permeability, nature of the stress applied, and duration of construction, designer usually considers the two conditions listed to determine which is more critical in the analysis a) At the end of construction, e.g. construction of river embankment in soft clay. Geotechnical

analysis maybe carried using total stress analysis with undrained shear strength parameters or effective stress analysis with drained shear strength parameters

b) Long-term e.g. construction of pervious reinforced earth retaining structure using free

draining backfill. Long-term geotechnical analysis is normally carried out using effective stress analysis with drained shear strength parameters and estimated or measured pore pressures.

Table 3.3 provided a more detail design conditions in relation to appropriate shear strengths for use in analyses of static loading conditions.


3-10 March 2009

Table 3.3 Design Conditions and Related Shear Strengths and Pore Pressures

Shear Strengths and Pore Pressures for Static Design Conditions Design Condition Shear Strength Pore Water Pressure During Construction and End-of-Construction

Free draining soils – use drained shear strengths related to effective stresses

Free draining soils – Pore water pressures can be estimated using analytical techniques such as hydrostatic pressure computations if there is no flow or using steady seepage analysis techniques (flow nets or finite element analyses).

Low permeability soils – use undrained shear strengths related to total stresses

Low-permeability soils = Total stresses are used, pore water pressures are set to zero in the slope stability computations.

Steady-State Seepage Conditions

Use drained shear strength related to effective stresses.

Pore water pressures from field measurements, hydrostatic pressure computations for no-flow conditions, or steady seepage analysis techniques (flow nets or finite difference analyses).

Sudden Drawdown Conditions

Free draining soils – use drained shear strengths related to effective stresses.

Free draining soils – First-stage computations (before drawdown) – steady seepage pore pressures as for steady seepage condition. Second- and third-stage computations (after drawdown) – pore water pressures estimated using same techniques as for steady seepage, except with lowered water level.

Low permeability soils – Three-

stage computations: First stage – use drained shear strength related to effective stresses, second stage – use undrained shear strengths related to consolidation pressures from the first stage, third stage – use drained strengths related to effective stresses, or undrained strengths related to consolidation pressures from the first stage, depending on which strength is lower – this will vary along the assumed shear surface.

Low-permeability soils – First-stage computations – steady state seepage pore pressures as described for steady seepage condition. Second–stage computations – total stresses are used, pore water pressures are set to zero. Third-stage computations – same pore pressures as free draining soils if drained strengths are used, pore water pressures are set to zero where undrained strengths are used.


March 2009 3-11

REFERENCES [1] Bishop A.V and Henkel D.J., The Measurement of Soil Properties in the Triaxial Test, E.Arnold, 1962. [2] Bowles, J.E. Foundation Analysis and Design. (Fourth edition). McGraw-Hill International, New York, 1992, 1004 p. [3] Brown, R.W., (1996) Practical foundation Engineering Handbooks, Mcgraw-Hill [4] BSI. Eurocode 7: Geotechnical Design – Part 1: General Rules (BS EN 1997-1 : 2004). British Standards Institution, London, 2004, 117 p. [5] Carter M. & Symons, M.V., Site Investigations and foundations Explained, Pentech Press, London [6] Donald P.Coduto, Foundation Design, Principles and Practices [7] Das, B.M., Principles of Geotechnical Engineering, PWK-Kent Publishing Company , Boston,MA., 1990 [8] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C., NAVFAC DM-7.1, May 1982, "Soil Mechanics" [9] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C.,NAVFAC DM-7.2, May 1982, "Foundations and Earth Structures" [10] Holtz, R.D., Kovacs, W.D. An Introduction to Geotechnical Engineering, Prentice-Hall, Inc. New Jersey [11] Koerner R.M . Construction and Geotechnical Method in Foundation Engineering, McGraw Hill, 1985. [12] Ladd C.C., Foott R., Ishihara K., Schlosser F., and Roulos H.G., Stress Deformation and Strength Characteristics, State of the Art Report, Session I, IX ICSMFE, Tokyo, Vol. 2, 1971, pp. 421 - 494. [13] Lambe T.W. and Whitman R.V., Soil Mechanics, John Wiley 8: Sons, 1969 [14] McCarthy D.J., Essentials of Soil Mechanics and Foundations. [15] Nayak N. V. I II Foundation Design Manual. Dhanpat Rai a Sons I 1982. [16] Peck R.B Hanson W.E. and Thornburn R.H., Foundation Engineering, John Wiley and Sons, 1974. [17] Smith C.N., Soil Mechanics for Civil and Mining Engineers. [18] Teng W.C., Foundation Design, Prentice Hall, 1984. [19] Terzaghi, K. & Peck, R.B. (1967). Soil Mechanics in Engineering Practice. (Second edition). Wiley, New York, 729 p.


3-12 March 2009

[20] U.S. Department of Transportation, Soil and Foundation, Reference Manual Volume 1 & 2 (2006)

CHAPTER 4 SOIL SETTLEMENT


March 2009 4-i

Table of Contents


List of Tables ......................................................................................................................... 4-ii


4 SOIL SETTLEMENT .............................................................................................................. 4-1

4.1 GENERAL CONCEPT .................................................................................................... 4-1

4.1.1 Immediate (Distortion) Settlement ................................................................ 4-1

4.1.2 Primary Consolidation ................................................................................... 4-2

4.1.3 Secondary Compression ................................................................................ 4-2

4.2 SETTLEMENT ON GRANULAR SOILS .............................................................. 4-2

4.3 ESTIMATION OF PRIMARY CONSOLIDATION IN COHESIVE SOIL .................................... 4-3

4.3.1 Normally Consolidated Soils .......................................................................... 4-5

4.3.2 Overconsolidated (Preconsolidated) Soils ....................................................... 4-5

4.3.3 Underconsolidated Soils ................................................................................ 4-6

4.4 RATE OF CONSOLIDATION .......................................................................................... 4-7

4.5 SECONDARY SETTLEMENT OF COHESIVE SOIL ............................................................. 4-9

4.6 DIFFERENTIAL SETTLEMENT ..................................................................................... 4-10

4.7 PLATE LOADING TEST FOR SETTLEMENT ESTIMATION ............................................... 4-12

4.8 SETTLEMENT OF RAFT/MAT FOUNDATIONS ............................................................... 4-12

REFERENCES ....................................................................................................................... 4-14


4-ii March 2009

List of Tables


4.1 Typical Allowable Total Settlements for Foundation Design 4-3

4.2 Typical Values of Tolerable Differential Settlement 4-11

List of Figures


4.1 Components Of Total Settlement Versus Log Time 4-1

4.2 Typical e – lop p Curve 4-4

4.3 Typical Consolidation Curve for Normally Consolidated Soil 4-5

4.4 Typical Consolidation Curve for Over Consolidated Soil 4-6

4.5 Typical Consolidation Curve for Under-Consolidated Soil 4-7

4.6 Average Degree of Consolidation U versus Time Factor, Tv under Various Drainage Conditions 4-8

4.7 Example 4.1 4-9

4.8 The Building was built partly on filled and partly on original ground, which resulted in cracks due to excessive differential settlement 4-10


March 2009 4-1

4 SOIL SETTLEMENT

4.1 GENERAL CONCEPT In geotechnical engineering, in particular foundation works for structures, engineers are interested in how much and how fast soil settlement will occur. Excessive settlement including (differential settlement) may cause structural damage as well as impair the functionality or serviceability of the structures. Soils whether cohesionless or cohesive, will experience settlements immediately after application of loads. Whether or not the settlements will continue with time after the application of the loads will be a function of how quickly the water can drain from the voids as explained in Item 3.2 Long-term consolidation-type settlements are generally not experienced in cohesionless soils where pore water can drain quickly or in dry or slightly moist cohesive soils where significant amounts of pore water are not present. Therefore, embankment settlements caused by consolidation of cohesionless or dry cohesive soil deposits are frequently ignored as they are much smaller compared to immediate settlements in such soils. The total soil settlement. St can be divided into 3 main components, namely immediate settlement, primary consolidation settlement, , and secondary compression settlement St = Si + Sc + Ss (4.1)

Si = immediate settlement Sc = primary consolidation settlement (time-dependent) Ss = secondary compression settlement

4.1.1 Immediate (Distortion) Settlement Immediate, or distortion, settlement (Si) occurs during application of load as excess pore pressure develops in the underlying soil. If the soil has a low permeability and it is relatively thick, the excess pore pressures are initially undrained. The foundation soil deforms due to the applied shear stresses with essentially no volume change, such that vertical compression is accompanied by lateral expansion. It should be recognized that most field evidence indicates that Si is usually not important design consideration especially in cohesive soils. It can usually be reduced by precompression or, to some extent, by a controlled loading program which allows consolidation to increase the soil stiffness and reduce the shear stress level in the foundation.

Ss

Sc

Si

Figure 4.1 Components Of Total Settlement Versus Log Time


4-2 March 2009

Immediate settlement although not actually elastic is usually estimated by using elastic theory, and the procedures for dealing with this problem can be found in textbooks on foundation engineering such as Soil and Foundation, FHWA and DID Geotechnical Guidelines. 4.1.2 Primary Consolidation Primary consolidation (Sc) develops with time as drainage allows excess pore pressure to dissipate. Volume changes, and thus settlement occur as stresses are transferred from the water (pore pressure) to the soil skeleton (effective stress). The rate of primary consolidation is governed by the rate of dissipation of pore water pressure. The estimation and rate of primary settlement in cohesive soil with low coefficient of permeability are dealt with in more details later in this Chapter. 4.1.3 Secondary Compression Secondary compression settlement (Ss) is the continuing, long term settlement which occurs after the excess pore pressures are essentially dissipated and after the effective stresses are practically constant. These further volume changes and increased settlements are due to drained creep, and are often characterized by a linear relationship between settlement and logarithm of time (refer Figure 4.1). Secondary compression is normally not very significant relative to the primary consolidation for inorganic clayey soil. However, for peats and highly inorganic soils, secondary compression constitutes a major part of the total settlement. Reader can refer to Holtz and Kovacs or Soil and Foundation, FHWA for guidance on the evaluation of secondary compression settlement. 4.2 SETTLEMENT ON GRANULAR SOILS Most methods for computing the primary settlements of foundations on granular soils are based on elastic theory or empirical correlations. Empirical correlations based on standard penetration test (SPT) generally provide an acceptable solution for predicting the settlement of a shallow foundation on granular soils. Poulos (2000) found that although soil behaviour is generally non-linear and highly dependent on effective stress level and stress history and hence should be accounted for in settlement analysis, the selection of geotechnical parameters, such as the shear and Young's modulus of soils, and site characterisation are more important than the choice of the method of analysis. Simple elasticity-based methods are capable of providing reasonable estimates of settlements. Based on elastic theory, the settlement, δf, of a shallow foundation can be calculated using an equation of the following general form:

δf=qnetBf'f

Es (4.2)

where qnet = mean net ground bearing pressure

Bf' = effective width of the foundation Es = Young’s modulus of soil f = a coefficient whose value depends on the shape and dimensions of the foundation,

the variation of soil stiffness with depth, the thickness of compressible strata, Poisson’s ratio, the distribution of ground bearing pressure and the point at which the settlement is calculated.


March 2009 4-3

Poulos & Davis (1974) gave a suite of elastic solutions for determining the coefficient 'f' for various load applications and stress distributions in soils and rocks. The increase of stress in soils due to foundation load can be calculated by assuming an angle of stress dispersion from the base of a shallow foundation. This angle may be approximated as a ratio of 2 (vertical) to 1 (horizontal) (Bowles, 1992; French, 1999). The settlement of the foundation can then be computed by calculating the vertical compressive strains caused by the stress increases in individual layers and summing the compression of the layers. A time correction factor has been proposed by Burland & Burbidge (1985) for the estimation of secondary settlement. Terzaghi et al (1991) also give an equation for estimating secondary settlement in a similar form. The commencement of secondary settlement is assumed to commence when the primary settlement completes, which is taken as the end of construction. 4.3 ESTIMATION OF PRIMARY CONSOLIDATION IN COHESIVE SOIL From the types of settlement described above, generally the most significant settlement is consolidation settlement. Consolidation settlement is time dependence. For low permeability soil with reasonably thickness, the primary consolidation may take very long time e.g., exceeding 10 years. Therefore, improvement method by shortening the consolidation process is essential to avoid distresses or failure due differential settlement after construction.

Table 4.1 Typical Allowable Total Settlements for Foundation Design

Type of Structure Typical Allowable Total Settlement, δa (in) (mm)

Office Buildings 0.5 – 2.1 (1.0 is the most common value)

12 – 50 (25 is the most common value)

Heavy Industrial Buildings 1.0 – 3.0 25 – 75 Bridges 2.0 50

(Source: Donald P.Coduto [19]) In general, lowering of the ground water table will leads to settlement of the ground. In fine-grained soils, prolonged lowering of water table will cause an increase in the effective stresses by extrusion of water from the voids leading to ground settlement. Primarily Consolidation, Sc (herein refer as ‘consolidation’) is a process when sudden application of a load to a saturated soil produces an immediate increase in pore water pressure. Over time, the excess pore water pressure will dissipate, the effective stress in the soil will increase and settlement will increase. Since shear strength is related to effective stress, it may be necessary to control the rate of construction to avoid a shear failure. The rate at which the excess water pressure dissipates, and settlement occurs, depends on the permeability of the soil, the amount of water to be expelled and the distance the water must travel (drainage path). The determination of consolidation is commonly based on the one-dimensional laboratory consolidation test results. Typically, the results are expressed in an e-log p plot which is the so-called “consolidation curve”, an example of which is as shown as in Figure 4.2. The followings parameters r may be obtained from the consolidation curve:

a) Initial void ratio, eo b) Compression index, Cc c) Recompression index, Cr d) Preconsolidation pressure, pc

4-4

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March 2009 4-5

4.3.1 Normally Consolidated Soils The settlement of a geotechnical feature or a structure resting on n layers of normally consolidated soils (pc = p ) can be computed from Figure 4.3 where n is the number of layers into which the consolidating yer is divided:

o

la

Sc = ∑ cc

1+e0

ni Ho log10

pf

po (4.4)

Figure 4.3 Typical Consolidation Curve for Normally Consolidated Soil The final effective vertical stress is computed by adding the stress change due to the applied load to the initial vertical effective stress. The total settlement will be the sum of the compressions of the n layers of soil. 4.3.2 Overconsolidated (Preconsolidated) Soils For overconsolidated clay, i.e., OCR >1, the soils could have in the past subjected to a greater stress than exists now. It maybe due to many factors including erosion of the weight of the natural soil deosit, removal of the weight of a previously placed fill or structures, etc. As a result of preconsolidation, the field state of stress will reside on the initially flat portion of the e-log p curve. Figure 4.4 illustrates the case where a load increment, ∆p, is added so that the final stress, pf. For this condition, the settlements for the case of n layers of overconsolidated soils will be computed by summing the settlements computed from each subdivided compressible layer within the zone of influence.

S = ∑ cc

1+e0

ni (cr log10

pcpo

+ cc log10 pfpc

)


4-6 March 2009

Figure 4.4 Typical Consolidation Curve for Over Consolidated Soil 4.3.3 Underconsolidated Soils When the state of effective stress of soils has not fully consolidated under an existing load, the soils is term as underconsolidation, i.e., OCR < 1. Consolidation settlement due to the existing load, will continue to occur under that load until primary consolidation is completed (i.e. under ∆po) even if no additional load is applied. This condition is represented in Figure 4.5. Thus, any additional load increment, ∆p, would have to be added to po. Consequently, if the soil is not recognized as being underconsolidated, the actual total primary settlement due to ∆po+∆p will be greater than the primary settlement computed for an additional load ∆p only, i.e., the settlement may be under-predicted. As a result of under-consolidation, the field state of stress will reside entirely on the virgin portion of the consolidation curve as shown in Figure 4.5.. The settlements for the case of n layers of under-consolidated soils are computed by Equation 4.5 that correspond to Figure 4.5.

S = ∑ Ho

1+eo n1 (cr log10

Pc

Po + cc log10

Pf

Pc (4.6)


March 2009 4-7

Figure 4.5 Typical Consolidation Curve for Under-Consolidated Soil 4.4 RATE OF CONSOLIDATION The average degree of consolidation, U at any time, t, can be defined as: U = St / Sult (4.7) Where St = Settlement at time of interest

Sult = Settlement at end of primary consolidation (i.e. at ultimate) when excess pore water pressures are zero throughout the consolidating layer

Figure 4.6 shows the average degree of consolidation (U) corresponding to a normalized time expressed in terms of a time factor, Tv, where :

Tv = cvt

Hd2 (4.8)

which can be written

t Tv Hd2

Cv (4.9)

cv = coefficient of consolidation (m

2/day)

Hd = The longest distance to a drainage boundary (m)

t = time (day)


4-8 March 2009

Figure 4.6 Average Degree of Consolidation U versus Time Factor, Tv under Various Drainage Conditions

Note that the longest drainage distance, Hd of a soil layer confined by more permeable layers on both ends is equal to one-half of the layer thickness. When confined by a more permeable layer on one side and an impermeable boundary on the other side, the longest drainage distance is equal to the layer thickness. The value of the dimensionless time factor Tv may be determined from Table 4.6 for any average degree of consolidation. U. The actual time, t, it takes for this percent of consolidation to occur is a function of the boundary drainage conditions, i.e., the longest distance to a drainage boundary, as indicated by Equation 4.8. By using the normalized time factor, Tv, settlement time can be computed for various percentages of settlement due to primary consolidation, to develop a predicted settlement-time curve. A typical settlement-time curve for a clay deposit under an embankment loading is shown in Figure 4.6 Coefficient of consolidation, cv can be obtained from laboratory consolidation test data. Two graphical procedures are commonly used for this i.e. the logarithm-of-time method (log t) proposed by Casagrande and Fadum (1940) and the square-root-of-time method proposed by Taylor (1948). These methods are can be found in various textbooks such as Holtz and Kovacs, and Soil and Foundations, FHWA.

U Tv 10 0.0077 20 0.0314 30 0.0707 40 0.126 50 0.196 60 0.286 70 0.403 80 0.567 90 0.848 100 Infinity

0

20

40

60

80

100

Perc

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atio

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4-10 March 2009

Mesri et al (1994) proposed correlating the secondary compression index, C , with the compression index, Cc, at the same vertical effective stress of a soil. He found that the C /Cc ration is the constant for a soil deposit (see Table 4.2). The time at which secondary consolidation is assumed to commence is not well defined. A pragmatic approach is to assume that the secondary consolidation settlement commences when 95% of the primary consolidation is reached (Terzaghi et al, 1991).

Table 4.2 Values of C /Cc for Geotechnical Materials

Material C /Cc

Granular soil

Shale and mudstone

Inorganic clays and silts

Organic clays and silts

Peat and muskeg

0.02 ± 0.01

0.03 ± 0.01

0.04 ± 0.01

0.05 ± 0.01

0.01 ± 0.01 (Source: Mesri et al [24])

4.6 DIFFERENTIAL SETTLEMENT Damage in structures due to settlement may be classified under 3 categories:

a) Architectural damage such as cracking in wall partitions and plaster b) Structural damage where the structural integrity are affected and c) Functional damage where the function of the structure may be impaired.

Figure 4.8 The Building was built partly on filled and partly on original ground, which resulted in cracks due to excessive differential settlement Normally, uniform settlement will not give rise to damage. It is the differential settlement that has to be controlled. However, differential settlement is difficult to estimate due especially to the non-homogeneity in the ground, and the large variations in the loadings between different supporting members. Figure 4.8 illustrates the appearance of crack due to differential settlement in a building. The limit of allowable settlement may be better expressed in terms of angular distortion, θ is θ =δ / L (4.11)


March 2009 4-11

Where δ = differential settlement in the structure L = horizontal distance between the 2 points where δ is considered. Skelton and McDonald established that for no architectural damage, θ must be less than 1/300 for buildings on individual footings. As a guide, reader can refer to Table 4.3 for the typical tolerable values of differential settlement.

Table 4.2 Typical Values of Tolerable Differential Settlement

Type of Structure Tolerable differential settlement, ß (radians) Comments

Circular steel petrol or fluid storage tanks: Fixed top Floating top

0.008 0.002 – 0.003

For floating top, value depends on details of top. Values apply to tanks on a flexible base. With rigid base slabs, such settlement will cause cracking and local buckling.

Tracks for overhead travelling crane. 0.003

Value taken longitudinally along track. Settlement between between tracks is not usually the controlling factor.

Rigid circular ring or mat footing for stacks, silos, water tanks etc.

0.002

Jointed rigid concrete pressure pipe. 0.015

Value is allowable angle change at joint. This is usually 2-4 times average slope of settlement profile. Damage to joint also depends on Longitudinal extension.

One- or two-storey steel framed warehouse with truss roof and flexible cladding.

0.006 – 0.008

Overhead crane, pipes, machinery or vehicles may limit tolerable values to less than this.

One- or two-storey houses or similar buildings with brick load-bearings walls.

0.002 – 0.003 Larger value is tolerable if most settlement has taken place before finishes are completed.

Structures with sensitive interior finishes such as plaster, ornamental stone or tiles.

0.001 -0.002

Multi-storey heavy concrete rigid framed structures on thick structural raft foundations.

0.0015

Damage to interior or exterior finish may limit value.

Span

Structure

Settlement profile

/3

Differential settlement

ß

(Source: Carter M, [7])


4-12 March 2009

4.7 PLATE LOADING TEST FOR SETTLEMENT ESTIMATION Guidelines and procedures for conducting plate loading tests are given in BS EN 1997-1:2004 (BSI, 2004) and DD ENV 1997-3:2000 (BSI, 2000b). The test should mainly be used to derive geotechnical parameters for predicting the settlement of a shallow foundation, such as the deformation modulus of soil. It may be necessary to carry out a series of tests at different levels. The plate loading test may also be used to determine the bearing capacity of the foundation in fine-grained soils, which is independent of the footing size. The elastic soil modulus can be determined using the following equation (BSI, 2000b):

Es= qnet b 1-vs

2

δp Is (4.12)

where

qnet = net ground bearing pressure δp = settlement of the test plate Is = shape factor b = width of the test plate νs = Poisson’s ratio of the soil Es = Young's modulus of soil

The method for extrapolating plate loading test results to estimate the settlement of a full-size footing on granular soils is not standardised. The method proposed by Terzaghi & Peck (1917) suggested the following approximate relationship in estimating the settlement for a full-size footing:

δf = δp2Bf

Bf+b

2 (4.13)

where: δp = settlement of a 30mm square test plate δf = settlement of foundation carrying the same bearing pressure Bf = width of the shallow foundation B = width of the test plate However, the method implies that the ratio of settlement of a shallow foundation to that of a test plate will not be greater than 4 for any size of shallow foundation and this could under estimate the foundation settlement. Bjerrum & Eggestad (1913) compared the results of plate loading tests with settlement observed in shallow foundations. They noted that the measured foundation settlement was much greater than that estimated from the method of Terzaghi & Pack (1917). Terzaghi et al (1991) also commented that the method is unreliable and is now recognized to be an unacceptable simplification of the complex phenomena. 4.8 SETTLEMENT OF RAFT/MAT FOUNDATIONS A raft/mat foundation is usually continuous in two directions and covers an area equal to or greater than the base area of the structure. A raft foundation is suitable when the underlying soils have a low bearing capacity or large differential settlements are anticipated. It is also suitable for ground containing pockets of loose and soft soils. In some instances, the raft foundation is designed as a cellular structure where deep hollow boxes are formed in the concrete slab. The advantage of a cellular raft is that it can reduce the overall weight of the foundation and consequently the net applied pressure on the ground. A cellular raft should be provided with sufficient stiffness to reduce differential settlement.

March 2

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4-14 March 2009

REFERENCES [1] Bishop A.V and Henkel D.J., The Measurement of Soil Properties in the Triaxial Test, E.Arnold, 1962. [2] Bowles, J.E. Foundation Analysis and Design. (Fourth edition). McGraw-Hill International, New York, 1992, 1004 p. [3] Brown, R.W., (1996) Practical foundation Engineering Handbooks, Mcgraw-Hill [4] BSI. Eurocode 7: Geotechnical Design – Part 1: General Rules (BS EN 1997-1 : 2004). British Standards Institution, London, 2004, 117 p. [5] Buisman, A.S.K. Results of long duration settlement tests. Proceedings of the First International Conference on Soil Mechanics and Foundation Engineering, Cambridge, Massachusetts, vol. 1, pp 103-101, 1931. [6] Burland, J.B. & Burbidge, M.C. Settlement of foundations on sand and gravel. Proceedings of Institution of Civil Engineers, Part 1, vol. 78, pp 1325-1381, 1985 [7] Carter M. & Symons, M.V., Site Investigations and Foundations Explained, Pentech Press, London [8] CGS, Canadian Foundation Engineering Manual, (Third edition). Canadian Geotechnical Society, Ottawa, 1992, 512 p. [9] Das, B.M., Principles of Geotechnical Engineering, PWK-Kent Publishing Company , Boston,MA., 1990 [10] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C., NAVFAC DM-7.1, May 1982, "Soil Mechanics" [11] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C.,NAVFAC DM-7.2, May 1982, Foundations and Earth Structures [12] Duncan, J.M. & Poulos, H.G. (1981). Modern techniques for the analysis of engineering problems in soft clay. Soft Clay Engineering, Elsevier, New York, pp 317-414. [13] DID Malaysia, Geotechnical Guidelines for D.I.D. works [14] EM 1110-2-1913. Design and Construction of Levees, U.S. Army Corp of Engineer, Washington, DC. [15] French, S.E. (1999). Design of Shallow Foundations, American Society for Civil Engineers Press, 374 p. [16] Foott R. and Ladd C.C., Undrained Settlement of Plastic and Organic Clays, Journal of Geotechnical Engineering Division, ASCE, Vol.107, No. GT8, August 1981. [17] ISE (1989). Soil-structure Interaction: The Real Behaviour of Structures. The Institution of Structural Engineers, London, 120 p. [18] Koerner R.M ., Construction and Geotechnical Method in Foundation Engineering, McGraw Hill, 1985.


March 2009 4-15

[19] Donald P.Coduto, Foundation Design, Principles and Practices [20] Ladd C.C., Foott R., Ishihara K., Schlosser F., and Roulos H.G., Stress Deformation and Strength Characteristics, State of the Art Report, Session I, IX ICSMFE, Tokyo, Vol. 2, 1971, pp. 421 - 494. [21] Lambe T.W. and Whitman R.V., Soil Mechanics, John Wiley 8: Sons, 1969 [22] Liao S.S.C. and Whitman R. V., Overburden Correction Factors for SPT' in Sand, Journal of the Geotechnical Engineering Division, ASCE. Vol. 112 No. 3, March 1986, pp. 373 - 377. [23] McCarthy D.J., Essentials of Soil Mechanics and Foundations. [24] Mesri G., discussion of New Design Procedure for stability of Soft Clays, by Charles C. Ladd and Roger Foott, Journal of the Geotechnical Engineering Division, ASCE, Vol.101, No. GT4. Froc. Paper 10664. April 1975. pp. 409 - 412. [25] Mesri, G., Lo, D.O.K. & Feng, T.W. (1994). Settlement of embankments on soft clays. Geotechnical Special Publication 40, American Society of Civil Engineers, vol. 1, pp 8-51. [26] Nayak N. V. I II Foundation Design Manual. Dhanpat Rai a Sons I 1982. [27] Parry, R.G. H. (1972). A direct method of estimating settlement in sands from SPT values. Proceedings of the Symposium on Interaction of Structures and Foundations, Midland Soil Mechanics and Foundation Engineering Society, Birmingham, pp 29-37. [28] Peck R.B Hanson W.E. and Thornburn R.H., Foundation Engineering, John Wiley and Sons, 1974. [29] Poulos, H.G. & Davis, E.H. (1974). Elastic Solutions for Soil and Rock Mechanics. John Wiley & Sons, New York, 411 p. [30] Poulos, H.G. (2000). Foundation Settlement Analysis – Practice versus Research. The Eighth Spencer J Buchanan Lecture, Texas, 34 p. [31] Poulos, H.G., Carter, J.P. & Small, J.C. (2002). Foundations and retaining structures – research and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. [32] Price, G. & Wardle, I.F. (1983). Recent developments in pile/soil instrumentation systems. Proceedings of the International Symposium on Field Measurements in Geomechanics, Zurich, vol. 1, pp 2.13-2.72. [33] Research and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. [34] Skempton A.W. and D.H. McDonald, "The Allowable Settlement of Buildings", Proc. Inst. Civil Eng., Vo1.5 Pt.3. 1956, pp. 727-784. [35] Skempton A.W., "The Bearing Capacity of Clays", Building Res. Congress, London Inst. Civ. Engrs., div.I:180, 1951. [36] Smith C.N., "Soil Mechanics for Civil and Mining Engineers".


4-16 March 2009

[37] Teng W.C., "Foundation Design", Prentice Hall, 1984. [38] Terzaghi, K. & Peck, R.B. (1967). Soil Mechanics in Engineering Practice. (Second edition). Wiley, New York, 729 p. [39] Thompson D.M. and Shuttler R.M., "Design of riprap slope protection against wind waves", Report 61, London, Construction Industry Research & Information Association. [40] Terzaghi, K. (1955). Evaluation of coefficients of subgrade reaction. Géotechnique, vol. 5, pp 297-321. [41] Tomlinson, M.J. (1994). Pile Design and Construction Practice. (Fourth edition). Spon, 411 p. [42] United Bureau States Department of the Interior, "Design of Small Dams” Bureau of Reclamation, Oxford and IBH Publishing Co., 1974. [43] Vesic, A.S. (1975). Bearing capacity of shallow foundations. Foundation Engineering Handbook, edited by Winterkorn, H.F. & Fang, H.Y., Van Nostrand Reinhold, New York, pp 121-147. [44] Zanen A., "Revetments", International Institute for Hydraulic and Environmental Engineering, Delft, Netherlands, 1978

CHAPTER 5 BEARING CAPACITY THEORY


March 2009 5-i

Table of Contents


List of Tables ......................................................................................................................... 5-ii


5.1 SHALLOW FOUNDATION ............................................................................................. 5-1

5.1.1 Bearing Capacity of Shallow Foundation ......................................................... 5-1

5.1.1.1 General ........................................................................................ 5-1

5.1.1.2 General Equation For Bearing Capacity ............................................ 5-2

5.1.2 Factors of Safety .......................................................................................... 5-5

5.1.3 Effects of Groundwater ................................................................................. 5-5

5.1.4 Foundation Near Crest of Slope ..................................................................... 5-6

REFERENCES ......................................................................................................................... 5-8


5-ii March 2009

List of Tables


5.1 Bearing Capacity Factors for Computing Ultimate Bearing Capacity of Shallow Foundations 5-4

List of Figures Figure Description Page

5.1 Generalized Loading and Geometric Parameter for a Spread Shallow Foundation 5-3

5.2 Groundwater Cases for Bearing Capacity Analysis 5-6

5.3 Linear Interpolation Procedures for Determining Ultimate Bearing Capacity of a Spread Shallow Foundation near the Crest of a Slope 5-7


March 2009 5-1

5 BEARING CAPACITY THEORY

5.1 SHALLOW FOUNDATION Shallow foundations, are generally more economical than deep foundations if they do not have to be installed deep into the ground and extensive ground improvement works are not required. They are often used to support structures at sites where ground are sufficiently strong. Unless a shallow foundation can be founded on strong rock, some noticeable settlement will occur. Design of shallow foundations should ensure that there is an adequate factor of safety against bearing failure of the ground, and that the settlements, including total and differential settlement, are limited to allowable values. For shallow foundations founded on granular soils, the allowable load is usually dictated by the allowable settlement, except where the ultimate bearing capacity is significantly affected by geological or geometric features. Examples of adverse geological and geometrical features are weak seams and sloping ground respectively. For shallow foundations founded on fine-grained soils, both the ultimate bearing capacity and settlements are important design considerations. High-rise structures or the presence of weak ground bearing materials do not necessarily stopping the design engineer from adopting shallow foundation system. Suitable design provision or ground improvement could be considered to overcome the difficulties. Some examples are given below: a. Design the foundations, structures and building services to accommodate the expected

differential and total settlements. b. Excavate weak materials and replace them with compacted fill materials. c. Carry out in-situ ground improvement works to improve the properties of the bearing materials.

Some of these methods are discussed in Chapter 9. d. Adopt specially designed shallow foundations, such as compensated rafts, to limit the net

foundation loads or reduce differential settlement.

5.1.1 Bearing Capacity of Shallow Foundation 5.1.1.1 General There are a many of methods for determining the bearing capacity of shallow foundations on soils. A preliminary estimate of allowable bearing pressure may be obtained on the basis of soil descriptions. Other methods include correlating bearing pressures with results of in-situ field tests, such as SPT N value and tip resistance of CPT. For example, Terzaghi & Peck (1917) proposed allowable bearing pressure of 10 N (kPa) and 5N (kPa) for non-cohesive soils in dry and submerged conditions respectively. This was based on limiting the settlement of footings of up to about 1 m wide to less than 25 mm, even if it is founded on soils with compressible sand pockets. Methods based on engineering principles can be used to compute the bearing capacity of soils and estimate the foundation settlement. This would require carrying out adequate ground investigation to characterize the site, obtaining samples for laboratory tests to obtain parameters and establishing a reliable model. Designs following this approach normally result in bearing pressures higher than the presumed allowable bearing pressures given in codes of practice.


5-2 March 2009

5.1.1.2 General Equation For Bearing Capacity Various equations have been established for calculating the bearing of shallow foundation. A comprehensive one which takes into consideration the shape of the foundation, inclination of loading, the base of the foundation and ground surface is as follows (GEO, 1993):

qu = Qu

f'Lf'B

c'Nc ζcs ζci ζct ζcg + 0.5 Bf' γs' Nγ ζγs ζγi ζγt ζγg + q Nq ζqs ζqi ζqt ζqg (5.1)

Where: Nc, Nγ, Nq = general bearing capacity factors which determine the capacity of a long strip

footing acting on the surface of a soil in a homogenous half space Qu = ultimate resistance against bearing capacity failure qu = ultimate bearing capacity of foundation q = overburden pressure at the level of foundation base c’ = effective cohesion of soil γs’ = effective unit weight of the soil Bf = least dimension of footing Lf = longer dimension of footing Bf’ = Bf – 2eB Lf’ = Lf – 2eL eL = eccentricity of load along L direction eB = eccentricity of load along B direction ζcs, ζγs, ζqs = influence factors for shape of shallow foundation ζci, ζγi, ζqi = influence factors for inclination road ζcg, ζγg, ζqg = influence factors for ground surface ζct, ζγt, ζqt = influence factors for tilting of foundation base

Figure 5.1 shows the generalized loading and geometric parameters for the design of a shallow foundation. The bearing capacity factors are given in Table 5.1. Equation 5.1 is applicable for the general shear type of failure of a shallow foundation, which is founded at a depth less than the foundation width. This failure mode is applicable to soils that are not highly compressible and have a certain shear strength, e.g. in dense sand. If the soils are highly compressible, e.g: in loose sands, punching failure may occur. Vesic (1975) recommended using a rigidity index of soil to define whether punching failure is likely to occur. In such case, the ultimate bearing capacity of the foundation can be evaluated based on Equation 5.1 with an additional set of influence factors for soil compressibility (Vesic,1975).

March 2

Fig

2009

gure 5.4 Genneralized Loa

Chapter 5 BEEARING CAPAC

ading and Ge

CITY THEORY

eometric Para

Y

ameter for a Spread Shallow Foundat

5-3

tion

5-4

Table 5

5.1 Bearing CCapacity Fact


tors for Com

CITY THEORY

puting Ultim

Y

ate Bearing Capacity of S

M

Shallow Foun

March 2009

ndations


March 2009 5-5

5.1.2 Factors of Safety The net allowable bearing pressure of a shallow foundation resting on soils is obtained by applying a factor of safety to the net ultimate bearing capacity i.e.

qu = q ult

F (5.2)

where

qult = ultimate net bearing capacity qu = allowable bearing capacity F = Factor of safety

The net ultimate bearing capacity should be taken as (qu – γ Df) where Df is the depth of soil above the base of the foundation and γ is the bulk unit weight of the soil. The selection of the appropriate factor of safety should consider factors such as:

(a) The frequency and likelihood of the applied loads (including different combination of dead load and live loads) reaching the maximum design level.

(b) Soil variability, e.g. soil profiles and shear strength parameters. The ground investigation helps increase the reliability of the site characterization.

(c) The importance of the structures and the consequences of their failures. In general, the minimum required factor of safety against bearing failure of a shallow foundation is in the range of 2.5 to 3.5. For most applications, a minimum factor of safety of 3.0 is adequate. Although the factor of safety is applied to the bearing capacity at failure, it is frequently used to limit the settlement of the foundation. 5.1.3 Effects of Groundwater The ultimate bearing capacity depends on the effective unit weight of the soil. Where groundwater is present, the effective stress and shear strength along failure plane will be smaller and the bearing capacity will be reduced. The effect of groundwater is accounted for by adjusting the γ s' in equation 5.1. and the three possible cases as shown in Figure 5.2 and describe below:

a) Case 1: Dw < D Use γ’ = γb = γ - γw where γb = weighted average buoyant unit weight

b) Case < + 2: D < Dw D B

Use ′ w 1- Dw-D

B

c) Case 3: D + B < Dw (no groundwater correction is necessary )

Use γ’ = γ


5-6 March 2009

Figure 5.5 Groundwater Cases for Bearing Capacity Analysis

5.1.4 Foundation Near Crest of Slope An approximate method is given in Geoguide 1: Guide to Retaining Wall Design (GEO HONG KONG, 1993) to determine the ultimate bearing capacity of a foundation near the crest of a slope. The ultimate bearing capacity can be obtained by linear interpolation between the value for the foundation resting at the edge of the slope and that at a distance of four times the foundation width from the crest. Equation 2.2 in section 2.2 can be used to estimate the ultimate bearing capacity for the foundation resting on the slope crest. Figure 5.3 summarises the procedures for the linear interpolation.

Lower Limit of Zone of influence

Dw D

Dw Dw D + B

Case 3 Case 1 Case 2

March 2

Figure

2009

e 5.6 Linear InterpolationSha


n Proceduresllow Foundat

CITY THEORY

s for Determition near the

Y

ning Ultimate Crest of a S

te Bearing CaSlope

apacity of a S

5-7

Spread


5-8 March 2009

REFERENCES [1] Bishop A.V and Henkel D.J., The Measurement of Soil Properties in the Triaxial Test, E.Arnold, 1962. [2] Bowles, J.E. Foundation Analysis and Design. (Fourth edition). McGraw-Hill International, New York, 1992, 1004 p. [3] Brown, R.W., (1996) Practical foundation Engineering Handbooks, Mcgraw-Hill [4] BSI. Eurocode 7: Geotechnical Design – Part 1: General Rules (BS EN 1997-1 : 2004). British Standards Institution, London, 2004, 117 p. [5] Buisman, A.S.K. Results of long duration settlement tests, Proceedings of the First International Conference on Soil Mechanics and Foundation Engineering, Cambridge, Massachusetts, vol. 1, pp 103-101, 1931. [6] Carter M. & Symons, M.V., Site Investigations and foundations Explained, Pentech Press, London [7] CGS, Canadian Foundation Engineering Manual, (Third edition). Canadian Geotechnical Society, Ottawa, 1992, 512 p. [8] Das, B.M., Principles of Geotechnical Engineering, PWK-Kent Publishing Company , Boston,MA., 1990 [9] DID Malaysia, Geotechnical Guidelines for D.I.D. works [10] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C., NAVFAC DM-7.1, May 1982, Soil Mechanics [11] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C.,NAVFAC DM-7.2, May 1982, Foundations and Earth Structures [12] EM 1110-2-1913. Design and Construction of Levees, U.S. Army Corp of Engineer, Washington, DC. [13] French, S.E. (1999). Design of Shallow Foundations, American Society for Civil Engineers Press, 374 p. [14] GCO (1990) Review of Design Method for Excavation, Geotechnical Control Office, Hong Kong [15] Hansen J.B . A Revised and Extended Formula for Bearing Capacity, Danish Geotechnical Institute, Bulletin No. 28; October 1968. [16] Holtz, R.D., Kovacs, W.D. An Introduction to Geotechnical Engineering, Prentice-Hall, Inc. New Jersey [17] ISE (1989). Soil-structure Interaction: The Real Behaviour of Structures. The Institution of Structural Engineers, London, 120 p. [18] Ladd C.C., Foott R., Ishihara K., Schlosser F., and Roulos H.G., "Stress Deformation and Strength Characteristics", State of the Art Report, Session I, IX ICSMFE, Tokyo, Vol. 2, 1971, pp. 421 - 494.


March 2009 5-9

[19] Lambe T.W. and Whitman R.V., Soil Mechanics, John Wiley 8: Sons, 1969 [20] Liao S.S.C. and Whitman R. V., Overburden Correction Factors for SPI' in Sand, Journal of the Geotechnical Engineering Division, ASCE. Vol. 112 No. 3, March 1986, pp. 373 - 377. [21] McCarthy D.J., Essentials of Soil Mechanics and Foundations. [22] Nayak N. V. I II Foundation Design Manual. Dhanpat Rai a Sons I 1982. [23] Peck R.B Hanson W.E. and Thornburn R.H., Foundation Engineering, John Wiley and Sons, 1974. [24] Poulos, H.G. & Davis, E.H. (1974). Elastic Solutions for Soil and Rock Mechanics. John Wiley & Sons, New York, 411 p. [25] Poulos, H.G., Carter, J.P. & Small, J.C. (2002). Foundations and retaining structures – research and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. [26] Research and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. [27] Skempton A.W., The Bearing Capacity of Clays, Building Res. Congress, London Inst. Civ. Engrs., div.I:180, 1951. [28] Smith C.N., Soil Mechanics for Civil and Mining Engineers. [29] Teng W.C., "Foundation Design", Prentice Hall, 1984. [30] Terzaghi, K. & Peck, R.B. (1967). Soil Mechanics in Engineering Practice. (Second edition). Wiley, New York, 729 p. [31] Terzaghi, K. (1955). Evaluation of coefficients of subgrade reaction. Géotechnique, vol. 5, pp 297-321. [32] United Bureau States Department of the Interior, "Design of Small Dams” Bureau of Reclamation, Oxford and IBH Publishing Co., 1974. [33] Vesic, A.S. (1975). Bearing capacity of shallow foundations. Foundation Engineering Handbook, edited by Winterkorn, H.F. & Fang, H.Y., Van Nostrand Reinhold, New York, pp 121-147.


5-10 March 2009


CHAPTER 6 SLOPE STABILITY


March 2009 6-i

Table of Contents

Table of Contents .................................................................................................................... 6-I

List of Tables ....................................................................................................................... 6-III

List of Figures ...................................................................................................................... 6-III

6.1 INTRODUCTION ..................................................................................................... 6-1

6.2 TYPE OF SLOPE INSTABILITIES ............................................................................... 6-1

6.2.1 Infinite Slope Failure .............................................................................. 6-1

6.2.2 Sliding Block Failure ............................................................................... 6-1

6.2.3 Circular Arc Failure ................................................................................. 6-2

6.3 GENERAL PROCEDURE FOR ANALYSIS ..................................................................... 6-3

6.3.1 Obtaining Subsurface Information ........................................................... 6-3

6.3.2 Determining of Soil Shear Strengths ........................................................ 6-3

6.3.3 Determining a Potential Slide Failure Surface ............................................ 6-3

6.4 PRINCIPLES OF ANALYSIS ...................................................................................... 6-4

6.4.1 Method of Analysis ................................................................................. 6-4

6.4.2 Stages of Stress Analysis ........................................................................ 6-4

6.4.2.1 Short-Term (or At-the-end-of-construction) .............................. 6-4

6.4.2.2 Long-term ............................................................................. 6-5

6.5 CIRCULAR ARC ANALYSIS ....................................................................................... 6-5

6.5.1 General Principles................................................................................... 6-5

6.5.2 Location of the Critical Slip Surface .......................................................... 6-6

6.5.4 Required Safety Factors .......................................................................... 6-7

6.5.5 Cut Slope in Clay .................................................................................... 6-7

6.5.6 Filled Slope/Embankment on Clay ............................................................ 6-8

6.5.7 Effects of Water ..................................................................................... 6-8

6.5.7.1 Effects on Cohesionless Soils ................................................... 6-9

6.5.7.2 Effects on Cohesive Soils ........................................................ 6-9

6.5.8 Method of Slides for Circular Failure ......................................................... 6-9

6.5.9 Finite Element Methods ........................................................................ 6-11

6.6 SLIDING BLOCK FAILURE ...................................................................................... 6-12

6.7 SLOPE STABILIZATION METHODS ......................................................................... 6-13

6.7.1 Slope Flattening ................................................................................... 6-13

6.7.2 Drainage ............................................................................................. 6-13

6.7.3 Buttressing or Counter Berm ................................................................. 6-14

6.7.4 Soil Nailing .......................................................................................... 6-14


6-ii March 2009

6.7.5 Geo-Synthetically Reinforcements .......................................................... 6-15

6.7.6 Retaining Walls .................................................................................... 6-15

REFERENCES ....................................................................................................................... 6-16

APPENDIX 6.A WORKED EXAMPLE: SLOPE STABILITY .................................................. 6A-1


March 2009 6-iii

List of Tables


6.1 Undrained Shear Strength and Consistency of Cohesive Soils (After Terzaghi & Peck and ASTM D2488-90) 6-5

6.2 Typical Drained Parameters For Effective Stress Analysis 6-5

6.3 Recommended Factors Of Safety 6-7

6.4 Guideline to Selection of Method of Slope Stability Analysis (After FHWA, Soils and Foundation Reference Manual) 6-11

6.5 Summary of Results 6A-2

List of Figures


6.1 Infinite Slope Failure 6-1

6.2 Sliding Block Failure Mechanism 6-2

6.3 Example of Circular Arc Failure Mechanism 6-2

6.4 Typical Circular Arc Failure Mechanism 6-6

6.5 Relationship Of Total Stress, Pore Pressure And Time 6-8

6.6 Effects Of Water Content On Cohesive Strength 6-9

6.7 Method of Slides 6-10

6.8 Geometric And Force Components For Sliding Block Analysis 6-12

6.9 Schematic View of Slope Regrading Work 6-13

6.10 Good Drainage System Critical to Stability of Slope 6-14

6.11 Butresses or Counter Berm for Slope Stabilsation 6-14

6.12 Typical Details of Soil Nail 6-15

6.13 Related Slope Configuration 6A-1

6.14 Stability Analysis of an Embankment Uses SLOPE/W Software 6A-3


6-iv March 2009


March 2

6.1 Slope sstrengtconstrupond) equipmconstruand che The facforces ttherefoa given 6.2 In gentypes, n 6.2.1 A slopeanalyzeof the s

6.2.2 Sliding layer of

2009

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March 2009 6-3

6.3 GENERAL PROCEDURE FOR ANALYSIS In general, analysis of slope stability would involves three basic parts:

a) Obtaining subsurface information b) Determining appropriate soil shear strengths and c) Determining a potential slide failure surface which provides the minimum safety factor

against failure under the various conditions

6.3.1 Obtaining Subsurface Information Previous works carried out at the site of interest generally can provide some subsurface information which are usually indicated in the design report or construction plans. The bore logs obtained may or may not be located close to the site and the engineer must determine if additional subsurface information is required. Additional boring(s) at the site are generally preferable. Other completed work in the nearby vicinity may also provide useful information. Soil type, thickness of each soil zone, depth to bedrock, and groundwater conditions must be known to proceed with a slope stability analysis. Reader can refer to Volume 6 Part 2 for further information on this matter. Before any analysis being carried out, it is always advisable to carry out geomorphological mapping of the project area. The observations during the mapping works can sometimes help significantly in deciding the types of tests, site investigation works and strengthening measures. The tell tale signs observed during the mapping works i.e., water seepages, ground saturation, erosion; mode of failure (deep seated or shallow slip) can be the references in the analysis and design stage. These geomorphologic features are always tie up with the estimation of the design parameter i.e., ground water condition, drainage adequacy and inherent properties (existence of discontinuities) which are difficult to retrieve from site investigation works. 6.3.2 Determining of Soil Shear Strengths The shear strength parameters of the embankment soil are normally defined in terms of a friction component (φ ) and a cohesion component (c). Shear strengths are usually determined from laboratory tests performed on specimens prepared by compaction in the laboratory or undisturbed samples obtained from exploratory soil borings. The laboratory test data may be supplemented with in situ field tests and correlations between shear strength parameters and other soil properties such as grain size, plasticity, and Standard Penetration Resistance (N) values. For a more detail discussion, reader can refer to Item. 3.3 of this Part. In general, for drained shear parameters for effective stress analysis, consolidated undrained (CU) can be used to obtained the effective soil strength parameter i.e., effective frictional angle φ‘ and effective cohesion c’. Shear box test can also be used in determining the strength parameter. The shear box sample shall be soaked in water for saturation and the shear rate shall be low to avoid misleading results. High cohesion (sometimes as high as 10kPa) and low frictional angle are the common error obtained from such tests if the saturation procedure is omitted. 6.3.3 Determining a Potential Slide Failure Surface All of the limit equilibrium methods require that a potential slip surface to be assumed in order to calculate the factor of safety. Circular slip surfaces can be assumed if the soil conditions are revealed to be relatively homogeneous. If the soil conditions are not homogeneous or if geologic anomalies appear, slope failures may occur on non-circular slip surfaces. The shape of the failure surface will depend on the problem geometry and stratigraphy, material characteristics (especially anisotropy), and the capabilities of the analysis procedure used. Commercially available computer


6-4 March 2009

programs such as SLOPE/W and STABL, which offer several analysis procedures, are useful for slope stability assessment.

6.4 PRINCIPLES OF ANALYSIS 6.4.1 Method of Analysis The methods for analysis of slope stability broadly used in engineering practice are limit equilibrium methods and finite element methods. The limit equilibrium method of slope stability analysis is used to evaluate the equilibrium of a soil mass tending to move down slope under the influence of gravity. A comparison is made between forces, moments, or stresses tending to cause instability of the mass, and those that resist instability. Two-dimensional (2-D) sections are analyzed and plane strain conditions are assumed. These methods assume that the shear strengths of the materials along the potential failure surface are governed by linear (Mohr-Coulomb) or nonlinear relationships between shear strength and the normal stress on the failure surface. Where estimates of movements as well as factor of safety are required to achieve design objectives, the effort required to perform finite element analysis can be justified. However, finite element analysis requires considerably more time and effort, compared to the limit equilibrium analysis and additional data related to stress-strain behavior of materials. Therefore, the use of finite element analysis is not justified for the sole purpose of calculating factors of safety. 6.4.2 Stages of Stress Analysis As mentioned in Para 3.3, shear strength of the soil varies with time. Thus, in slope stability analysis, it is important for the designer to understand and determine at which point in time i.e. before, during or after construction that is more critical and yield the lowest factor of safety. Generally, the two conditions considered are: 6.4.2.1 Short-Term (or At-the-end-of-construction) Analyses of the short-term condition of stability are normally performed in terms of total stress (using undrained shear strength parameters), with the assumption that any pore water pressure set up by the construction activity will not dissipate at all. However, in some construction works such as large earth dams or embankments, the construction period is relatively long, and some dissipation of the excess pore water pressure is likely. Under these conditions, a total stress analysis would yield a value of factor of safety on the low side, possibly resulting in un-economic design. For undrained shear strength of saturated soil, φ can be assumed as zero and knowledge of the pore water pressure (i.e. the phreatic line) is not necessary since total stress can be expressed independently of effective stress at failure. For instance, the total stress analysis must be used for the construction of coastal bund in soft clay and it usually gives the worst critical factor of safety. Unconsolidated Undrained (UU) Triaxial test is usually used to obtain the undrained strength parameter of the soil. Extra care shall be given during the test when the soil samples are not fully saturated. For soft to very soft clay such as coastal alluvium clay, in-situ strength test using in-situ vane shear test should be used to determine the undrained shear strength. Typical values of undrained shear strength for Malaysia coastal alluvium clay ranges from 10 to 20 kPa. Table 6.1 gives some typical values of undrained shear strength, c which may be used for preliminary analysis or to check laboratory test results


March 2009 6-5

Table 6.1 Undrained Shear Strength and Consistency of Cohesive Soils

Consistency Undrained Shear Strength, Su (kPa) Visual Identification

Very soft < 12 Thumb can penetrate more than 25 mm Soft 12 – 25 Thumb can penetrate about 25 mm

Medium 25 -50 Thumb can penetrate with moderate effort

Stiff 50 – 100 Thumb will indent soil about 8 mm

Very stiff 100 – 200 Thumb will not indent but readily indent with thumbnail

(After Terzaghi & Peck and ASTM D2488-90) 6.4.2.2 Long-term Long-term stability analysis is normally carried out using effective stress analysis with drained shear strength parameters. For cohesive or clayey soil, total stress analysis (for short-term) in addition to the effective stress analysis (for long-term) are carried out to determine the most critical factor of safety. As granular or sandy soils are more permeable than cohesive or clayey soils, drainage of excess pore pressure in sandy soil occurs much more rapidly. Hence, only effective stress analysis is usually required. Effective stress analysis requires the estimation of the drained strength parameters c’, φ’ and pore pressures. For pure free draining sands, φ = φ’ and c = 0. Under conditions of steady seepage, the phreatic line can be obtained from the flow net.

Some common drained strength parameters, φ' and c’ adopted in the slope analysis are as follows:-

Table 6.2 Typical Drained Parameters For Effective Stress Analysis

Soil type Effective friction angle φ‘ Effective cohesion c’

Well compacted soil 28o – 30o 2 – 5 kPa Residual soil grade V to VI 30o – 32o 5 – 10kPa Residual soil grade IV to V 32o – 35o 10 – 15kPa Note:- • The values above are just for references. Test shall be carried out before any

analysis is carried out. It is advisable to limit the cohesion to not more than 15kPa even with lab test results. The cohesion shows in test are sometimes apparent and the changes are subjected to external factors i.e., weathering process etc

• Description of grade of residual soil: Grade VI = residual soil : Grade V = completely weathered rock ; Grade IV =

highly weathered 6.5 CIRCULAR ARC ANALYSIS 6.5.1 General Principles Figure 6.4 shows a potential slide mass defined by a predetermined circular arc slip surface. If the shear resistance of the soil along the slip surface exceeds that necessary to provide equilibrium, the mass is stable. If the shear resistance is insufficient, the mass is unstable. Thus, the stability or instability of the mass depends on its weight, the external forces acting on it, the shear strengths


6-6 March 2009

and pore-water pressures along the slip surface. Circular arc slip surface is often used because it simplifies the calculations by just conveniently summing up the moments or forces about the center of the circle. Also, circular slip surfaces are generally sufficient for analyzing relatively homogeneous embankments or slopes.

Figure 6.4 Typical Circular Arc Failure Mechanism

The requirement for static equilibrium of the soil mass are used to compute a factor of safety with respect to shear strength. The factor of safety is defined as the ratio of the available shear resistance to the driving force that can cause movement of the slope. In Figure 6.4, the factor of safety (FOS) is

FOS = Resisting MomentDriving Moment

= Total shear strength x Ls

Weight force × Lw (6.1)

Limit equilibrium analysis assumes the factor of safety is the same along the entire slip surface. A value of factor of safety greater than 1.0 indicates that shear resistance exceeds the required for equilibrium and that the slope will be stable with respect to sliding along the assumed particular slip surface analyzed. A value of factor of safety less than 1.0 indicates that the slope will be unstable. 6.5.2 Location of the Critical Slip Surface The critical slip surface is defined as the surface with the lowest factor of safety. Because different methods of analysis like Bishop’s, Janbu’s and Spencer’s adopt different assumptions, the location of the critical slip surface can vary among different methods of analysis. The critical slip surface for a given problem analyzed by a given method is found by a systematic procedure of generating trial slip surfaces until the one with the minimum factor of safety is obtained. Searching schemes may vary with the assumed shape of the slip surface and the computer program used. All external loadings imposed on the embankment or ground surface should be represented in slope stability analysis, including loads imposed by water pressures, structures, surcharge loads, anchor forces, or other causes.

Fill Surface after Failure

Fill Weight Force

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CenterLw

Ls


March 2009 6-7

6.5.4 Required Safety Factors Appropriate factors of safety are required to ensure adequate performance of embankments throughout their design lives. Two of the most important considerations that determine appropriate magnitudes for factor of safety are uncertainties in the conditions being analyzed, including shear strengths and consequences of failure (both economic loss and loss of life) or unacceptable performance. The values of factor of safety listed in Table 6.3 provide a guidance and are not prescribed for slopes of embankment dams. Higher or lower values might be warranted in respect of the degree of uncertainties in the conditions being analyzed, economic loss and loss of life.

Table 6.3 Recommended Factors Of Safety

6.5.5 Cut Slope in Clay For cut slope, the effective stress reduces with time owing to the stress relief after removal of load. This reduction will allow the clay to expand and absorb water, which will lead to a decrease in the clay strength with time. For this reason, the factor of safety of a cut slope in clay may decrease with time. Cut slopes in clay should be designed by using effective strength parameters and the effective stresses that will exist in the soil after the pore pressures have come into equilibrium under steady seepage condition. These changes in the values of total stress and pore pressure with time are shown here in Figure 6.5(a).

Type of slopes End of construction (short-term)

Long-term (steady-stage seepage)

Rapid drawdown 3

1. Embankment and Natural Slope 1 1.3 1.4 1.1 – 1.2 4

2. Cut or Excavated Slope 2 1.3 1.4 1.1 - 1.2 4

Notes 1. Applicable to filling for river bank, water retention facilities, levees, sea wall, stockpiles, earth

retaining works. It also includes natural slopes such as river bank and valley slopes. 2. Applicable to excavated slope including foundation excavation, excavated river and retention

facilities, sea wall and other earth retaining works. 3. Rapid drawdown occurs when it is assumed that drawdown is very fast, and no drainage

occurs in materials with low permeability; thus the term “sudden” drawdown. 4. For submerged or partially submerged slopes, the possibility of low water events and rapid

drawdown should be considered. FOS of 1.1 to 1.2 for rapid drawdown recommended here are for cases where rapid drawdown represents an infrequent loading condition. In cases where rapid drawdown represents a frequent loading condition, as in river bank subjected fluctuations in water level and pumped storage projects, the factor of safety should be higher.


6-8 March 2009

During slope cutting, frequent inspections and mapping shall be carried out by experience geologist to ensure no adverse “inherent” geological features i.e., soil bedding, relicts and rock discontinuities (if rock cutting). If these adverse features are found on slope outcrop, strengthening measures such as soil nailing can be specified to improve the stability of the slope. Horizontal drains can be installed at areas where water seepages are found during cutting to lower the ground water table. Always avoid cutting slope with large catchment behind the slope. Area with large catchment always associated with high ground water table. If it is unavoidable, Horizontal drains and deep trench drains shall be included in the design to lower the ground water table

6.5.6 Filled Slope/Embankment on Clay Excess pore water pressures are created when fills are placed on clay or silt. Provided the applied loads do not cause the undrained shear strength of the clay or silt to be exceeded, as the excess pore water pressure dissipates consolidation will occur, and the shear strength increases with time as illustrated in Figure 6.5(b). For this reason, the factor of safety increases with time under the load of the fill. Hence, the most critical state for the stability of an filled embankment is normally the short-term or end-of-construction condition where total stress analysis with undrained shear parameters are required. 6.5.7 Effects of Water Besides gravity, water (both surface and ground water) is a major factor in slope instability. In addition, ground water table induced failure is always deep seated and catastrophic. Ground water table is one of the most difficult parameter to be assumed or estimated. Hence, if necessary standpipes or piezometers can be installed to monitori and ascertain the fluatuation and worst ground water levels to be used either in design or verification of design. If the slope is subjected to inundation and changes in the water levels such as dam, pond, or river subjected to tidal effects, the designer should consider the possible effects of rapid draw down of water levels in the stability analysis. For rapid drawdown analysis of soils with low permeability (less than 10-4

cm/sec), it is assumed that the drop in water level is so fast that no drainage can occur in the soil. For this prupose, drained strengths with appropriate phreatic line are used for stability analysis.

Excavation/cut Time

Increase in pore pressure

σ’ u σ

σ’

Construction/fill Time

decrease in pore pressure u

σ

a

b

Figure 6.5 Relationship Of Total Stress, Pore Pressure And Time


March 2009 6-9

Instability of natural slopes is often related to high internal water pressures associated with wet weather periods. It is appropriate to analyze such conditions as long-term, steady-state seepage conditions, using drained strengths and the highest probable position of the piezometric surface within the slope. 6.5.7.1 Effects on Cohesionless Soils In cohesionless soils, water does not affect the angle of internal friction (φ ’). The effect of water on cohesionless soils below the water table is to decrease the intergranular stress between soil grains (efffective normal stress, σn'), which decreases the frictional shearing resistance. 6.5.7.2 Effects on Cohesive Soils An increase in absorbed moisture is a major factor in the decrease in strength of cohesive soils as shown schematically in Figure 6.6. Water absorbed by clay minerals causes increased water contents that decrease the cohesion of clayey soils. These effects are amplified if the clay mineral happens to be expansive, e.g., montmorillonite. Some weak rocks such as shales, claystones, and siltstones tend to disintegrate into a clay soil if water is allowed to percolate into them. This transformation from rock to clay often leads to settlement and/or shear failure of the slope.

6.5.8 Method of Slides for Circular Failure For slope stability analysis, the method of dividing the soil mass into vertical slides is most commonly used and illustrated in Figure 6.5 (a). The forces acting on each slide is shown in Figure 6.7 (b)

cohesive strength

water content

Figure 6.6 Effects Of Water Content On Cohesive Strength


6-10 March 2009

Figure 6.7 Method of Slides Fellenius’s method of slides is one of the oldest methods used. Subsequently, several other methods basing on the method of slides were developed which include Bishop’s Simplified Method, Janbu’s Simplified Method, Morgenstern and Price’s Method and Spencer’s Method. Fellenius’s method is normally more conservative and gives unrealistically lower factors of safety than other more refined methods. The only reason this method is discussed here is to demonstrate the basic principles of slope stability. Reader can refer to Appendix A Example A.1 on the application of Fellenius’s Method of slides in deriving the factor of safety. Various methods may result in different values of factor of safety because:

(a) the various methods employ different assumptions to make the problem statically determinate

(b) some of the methods do not satisfy all conditions of equilibrium.

(b) Forces on a slide with effect of water

(a) Method of Slides


March 2009 6-11

Table 6.4 Guideline to Selection of Method of Slope Stability Analysis (After FHWA, Soils and

Foundation Reference Manual)

Foundation Soil Type

Type of Analysis Source of Strength Parameters Remarks (see Note 1)

Cohesive

Short-term or end of construction

• UU or field vane shear test or CU triaxial test.

• Undrained strength parameters tested at p0

(ground overburden stress)

Use Bishop Method. An angle of internal friction should not be used to represent an increase of shear strength with depth.

Stage Use Bishop Method at each construction (embankments on soft clays – build embankment in stages with waiting periods to take advantage of clay strength gain due to consolidation

• CU triaxial test. Some samples should be consolidated to higher than existing in-situ stress to determine clay strength gain due to consolidation under staged fill heights.

• Use undrained strength parameters at appropriate p0 for staged height

stage of embankment height. Consider that clay shear strength will increase with consolidation under each stage. Consolidation test data needed to estimate length of waiting periods between embankment stages. Piezometers and settlement devices should be used to monitor pore water pressure dissipation and consolidation during construction

Long-term • CU triaxial test with pore water pressure measurements or CD triaxial test.

• Use effective strength parameters.

Use Bishop Method with (embankment combination of cohesion and on soft clays angle of internal friction and clay cut (effective strength parameters slopes. from laboratory test).

Existing failure planes

• Direct shear or direct simple shear test. Slow strain rate and large deflection needed.

• Use residual strength parameters.

Use Bishop, Janbu or Spencer Method to duplicate previous shear surface.

Granular All types

• Obtain effective friction angle from charts of standard penetration resistance (SPT) versus friction angle or from direct shear tests.

Use Bishop Method with an effective stress analysis.

Note 1: Methods recommended represent minimum requirement. More rigorous methods such as Spencer’s method should be used when a computer program has such capabilities.

6.5.9 Finite Element Methods The finite element methods can be used to compute stresses and displacements in earth structures caused by applied loads. The method is particularly useful for soil-structure interaction problems, in which structural members interact with a soil mass. The stability of a slope cannot be determined directly from finite element analysis, but the computed stresses in a slope can be used to compute a factor of safety. Use of the finite element methods for stability problems is a complex and time-consuming process.


6-12 March 2009

Finite element analysis can provide estimates of displacements and construction pore water pressures. This is useful for the field control of construction works, or when there is concern for damage to adjacent structures. If the displacements and pore water pressures measured in the field differ greatly from those computed, the reason for the difference should be investigated. Finite element analysis provides displacement pattern which may show potential and possibly complex failure mechanisms. The validity of the factor of safety obtained from limit equilibrium analysis depends on locating the most critical potential slip surfaces. In complex conditions, it is often difficult to anticipate failure modes, particularly if reinforcement or structural members such as geotextiles, concrete retaining walls, or sheet piles are included. Once a potential failure mechanism is recognized, the factor of safety against a shear failure developing by that mode can be computed using conventional limit equilibrium procedures. Finite element analysis provides estimates of mobilized stresses and forces. The finite element method may be particularly useful in judging what strengths should be used when materials have very dissimilar stress-strain and strength properties, i.e., where strain compatibility is an issue. The finite element methods can help to identify local regions where “overstress” may occur and cause cracking in brittle and strain softening materials. 6.6 SLIDING BLOCK FAILURE Block slide failure mechanisms are defined by dividing into straight line segments defining an active wedge, central block, and passive wedge. An example of the wedge is shown in Figure 6.8

Figure 6.8 Geometric And Force Components For Sliding Block Analysis


March 2009 6-13

The factor of safety for the wedge can be and computed by:

FOS = Horizontal Resisting Forces Horizontal Driving forces =

Pp + cL

Pa (6.2)

Pa = Active force (driving) Pp = Passive force (resisting) cL = Resisting force due to cohesive clay

For method of computation of the active force and passive forces reader can refer to the Chapter 7 on retaining wall. 6.7 SLOPE STABILIZATION METHODS Slope stabilization methods generally aim to reduce driving forces, increase resisting forces, or both. Driving forces can be reduced by excavation of materials from appropriate part of the unsuitable ground and drainage of water to reduce the hydrostatic pressures acting on the unstable zone. Resisting forces can be increased by introducing soil reinforcements, such as soil nails and geo-synthetic materials, and retaining structures or other supports. 6.7.1 Slope Flattening Slope flattening is a common method for increasing the stability of a slope by reducing the driving forces that contribute to movements. Often, it is the first option to be considered when stabilizing a slope.

Regrading Slope Profile

Existing Slope Profile

Figure 6.9 Schematic View of Slope Regrading Work

6.7.2 Drainage Surface (berm, toe, interceptor, and cascade drains) and subsurface (horizontal drains and gravel trenches) drainages are essential for treatment of any slide or potential slide. Proper drainage system can reduce the destabilizing hydrostatic and seepage forces on a slope as well as the risk of erosion.

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6-16 March 2009

REFERENCES [1] Bishop A.V and Henkel D.J., The Measurement of Soil Properties in the Triaxial Test, E.Arnold, 1962. [2] Bowles, J.E. Foundation Analysis and Design. (Fourth edition). McGraw-Hill International, New York, 1992, 1004 p. [3] Brown, R.W., (1996) Practical foundation Engineering Handbooks, Mcgraw-Hill [4] BSI. Eurocode 7: Geotechnical Design – Part 1: General Rules (BS EN 1997-1 : 2004). British Standards Institution, London, 2004, 117 p. [5] Carter M. & Symons, M.V., Site Investigations and foundations Explained, Pentech Press, London [6] CGS, “Canadian Foundation Engineering Manual”, (Third edition). Canadian Geotechnical Society, Ottawa, 1992, 512 p. [7] Das, B.M., Principles of Geotechnical Engineering, PWK-Kent Publishing Company , Boston,MA., 1990 [8] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C., NAVFAC DM-7.1, May 1982, Soil Mechanics [9] DID Malaysia, Geotechnical Guidelines for D.I.D. works [10] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C.,NAVFAC DM-7.2, May 1982, Foundations and Earth Structures [11] Duncan, J. M., Buchignani, A. L., and DWet, M., An Engineering Manual for Slope Stability Studies, Department of Civil Engineering, Geotechnical Engineering, Virginia Polytechnic Institute and State University, Blacksburg, VA, 1987. [12] Duncan, J.M. & Poulos, H.G. (1981). Modern techniques for the analysis of engineering problems in soft clay. Soft Clay Engineering, Elsevier, New York, pp 317-414. [13] EM 1110-2-1902. Engineering and Design of Slope Stability, U.S. Army Corp of Engineer, [14] GCO (1984). Geotechnical Manual for Slope”. (Second Edition). Geotechnical Control Office, Hong Kong [15] GCO (1990) Review of Design Method for Excavation, Geotechnical Control Office, Hong Kong [16] Holtz, R.D., Kovacs, W.D. An Introduction to Geotechnical Engineering, Prentice-Hall, Inc. New Jersey [17] Huang Y.H., Stability Analysis of Earth Slopes, Van Nostrand Reinhold, 1983. [18] Ladd C.C., Foott R., Ishihara K., Schlosser F., and Roulos H.G., "Stress Deformation and Strength Characteristics", State of the Art Report, Session I, IX ICSMFE, Tokyo, Vol. 2, 1971, pp. 421 - 494.


March 2009 6-17

[19] Lambe T.W. and Whitman R.V., "Soil Mechanics", John Wiley 8: Sons, 1969 [20] McCarthy D.J., "Essentials of Soil Mechanics and Foundations". [21] Mesri G., discussion of "New Design Procedure for stability of Soft Clays". by Charles C. Ladd and Roger Foott, Journal of the Geotechnical Engineering Division, ASCE, Vol.101, No. GT4. Froc. Paper 10664. April 1975. pp. 409 - 412. [22] Mesri, G., Lo, D.O.K. & Feng, T.W. (1994). Settlement of embankments on soft clays. Geotechnical Special Publication 40, American Society of Civil Engineers, vol. 1, pp 8-51. [23] Nakashima, E., Tabara, K. & Maeda, Y.C. (1985). Theory and design of foundations on slopes. Proceedings of Japan Society of Civil Engineers, no. 355, pp 41-52. (In Japanese). [24] Parry, R.G. H. (1972). A direct method of estimating settlement in sands from SPT values. Proceedings of the Symposium on Interaction of Structures and Foundations, Midland Soil Mechanics and Foundation Engineering Society, Birmingham, pp 29-37. [25] Peck R.B Hanson W.E. and Thornburn R.H., “Foundation Engineering", John Wiley and Sons, 1974. [26] Poulos, H.G., Carter, J.P. & Small, J.C. (2002). Foundations and retaining structures – research and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. [27] Price, G. & Wardle, I.F. (1983). Recent developments in pile/soil instrumentation systems. Proceedings of the International Symposium on Field Measurements in Geomechanics, Zurich, vol. 1, pp 2.13-2.72. [28] Research and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. [29] Skempton A.W. and D.H. McDonald, "The Allowable Settlement of Buildings", Proc. Inst. Civil Eng., Vo1.5 Pt.3. 1956, pp. 727-784. [30] Smith C.N., "Soil Mechanics for Civil and Mining Engineers". [31] Teng W.C., "Foundation Design", Prentice Hall, 1984. [32] Terzaghi, K. & Peck, R.B. (1967). Soil Mechanics in Engineering Practice. (Second edition). Wiley, New York, 729 p. [33] United Bureau States Department of the Interior, "Design of Small Dams” Bureau of Reclamation, Oxford and IBH Publishing Co., 1974. [34] Huang Y.H., Stability Analysis of Earth Slopes, Van Nostrand Reinhold, 1983.


6-18 March 2009



March 2009 6A-1

APPENDIX 6.A WORKED EXAMPLE: SLOPE STABILITY A.1 Problem The worked example presented herein illustrates the application of stability analysis by way of the Fellenius method of slices to determine the factor of safety in terms of effective stresses. The related slope configuration is shown in Figure 6.13 below.

Figure 6.13 Related Slope Configuration The applicable soil properties and strength parameters are given as follows:

i. ii. iii. iv.

Soil unit weight (above & below water table), γs Effective cohesion, c’ Effective angle of shearing resistance, φ’ The soil mass is divided into slices of 1.5m wide sing theexpression below (Eqn. 3.1), the resulting factor ofsafety is established as follows.

= 20 kN/m3

= 10 kN/m2

= 29°

F = c'La+ tan ' ∑ W cosα-ul∑ W sin α

(6.3)

Solution:

i. The weight of each slice, W = γsbh = 20 x 1.5 x h = 30h kN/m


6A-2 March 2009

ii. The height of each slice is set off below the centre of the base, and the

normal and tangential components, h cos α and h sin α respectively are determined graphically as shown in Figure 3.3. Thus:

W cos α W sin α

==

30h cos α 30h sin α

iii. The pore water pressure at the centre of the base of each slice is taken to be

γwzw, where zw is the vertical distance of the centre point below the water table (Fig 3.3 refers). [Note: This procedure slightly overestimates the pore water pressure, which strictly should be γwze, where ze is the vertical distance below the point of intersection of the water table and the equipotential through the centre of the slice base. The error involved is however, on the safe side].

iv. From Figure 6.13, the overall arc length, La is calculated as 14.35m. v.

The results are summarised in Table 6.2 below.

Table 6.5 Summary of Results

Slice No. h cos α (m)

h sin α (m)

u (kN/m2)

l (m)

u.l (kN/m)

1 2 3 4 5 6 7 8

0.75 1.80 2.70 3.25 3.45 3.10 1.90 0.55

- 0.15 - 0.10 0.40 1.00 1.75 2.35 2.25 0.95

5.9 11.8 11.2 18.1 17.1 11.3

0 0

1.55 1.50 1.55 1.10 1.70 1.95 2.35 2.15

9.1 17.7 25.1 29.0 29.1 22.0

0 0

17.50 8.45 14.35 132.0

vi. Hence:

Σ W cos α = 30 x 17.50 Σ W sin α = 30 x 8.45 Σ (W cos α - ul) = 525 – 132

= 525 kN/m = 254 kN/m = 393 kN/m

F = c'La+tan ' ∑ Wcosα-ul

∑ Wsin α

= 10x14.35 +(0.554x393)

254

= 143.5+218

254

= 1.42


March 2009 6A-3

A.2 PROBLEM Figure 6.14 shows a slope stability analysis of an embankment on soft clay using a commercial software; SLOPE/W. The soil stratums are as illustrated in Figure 6.14. In order to increase the factor of safety, two layers of high strength geotextiles were adopted. For embankment on soft soils, undrained condition is adopted.

Figure 6.14 Stability Analysis of an Embankment Uses SLOPE/W Software


6A-4 March 2009


CHAPTER 7 RETAINING WALL


March 2009 7-i

Table of Contents

Table of Contents…………………………………………………………………………………………...…7-i

List of Tables ........................................................................................................................ 7-II

List of Figures ....................................................................................................................... 7-II

7.1 GENERAL ..................................................................................................................... 7-1

7.2 TYPE OF RETAINING WALLS .......................................................................................... 7-1

7.3 SHEAR STRENGTH – LATERAL EARTH PRESSURE RELATIONSHIP ..................................... 7-2

7.4 LATERAL EARTH PRESSURE ........................................................................................... 7-4

7.4.1 At-Rest Lateral Earth Pressure .......................................................................... 7-4

7.4.2 Active and Passive Lateral Earth Pressures ........................................................ 7-5

7.4.2.1 Rankine’s Theory ............................................................................ 7-5

7.4.2.2 Coulomb’s Theory ........................................................................... 7-8

7.4.2.3 Effects of Wall Friction ..................................................................... 7-9

7.4.3 Lateral Earth Pressure Due to Ground Water ................................................... 7-14

7.4.4 Lateral Pressure from Surchage ...................................................................... 7-14

7.5 STABILITY OF RIGID RETAINING WALL ....................................................................... 7-17

7.5.1 Sliding/Translational Stability ......................................................................... 7-19

7.5.2 Overturning Stability...................................................................................... 7-19

7.5.3 Bearing Capacity Failure ................................................................................ 7-20

7.5.4 Global Stability .............................................................................................. 7-20

7.5.5 Selection of Backfill Materials ......................................................................... 7-21

7.5.6 Design Wall Drainage System ......................................................................... 7-21

7.5.7 Design Example – Gravity/Cantilever Reinforced Concrete Wall ......................... 7-23

7.6 FLEXIBLE WALL SYSTEM ............................................................................................. 7-25

7.6.1 General ........................................................................................................ 7-25

7.6.2 Types of Flexible Walls .................................................................................. 7-26

7.6.3 Sheet Pile Wall .............................................................................................. 7-27

7.6.3.3 Design of Anchor - General ............................................................ 7-30

7.6.3.4 Some Considerations on Sheet Pile Wall Design ............................... 7-31

7.6.3.3 Cantilever Steel Sheet Pile Retaining Wall - Example ....................... 7-33

REFERENCES........................................................................................................................ 7-38


7-ii March 2009

List of Tables


7.1 Wall Displacements Required to Develop Active and Passive Earth Pressures (Wu, 1975) 7-5

7.3 Calculation Table 7-24

7.4 Permissible Steel Stress of Sheet Pile 7-32


7.1 Forces Acting On Retaining Wall And Common Terminology 7-1

7.2 Type of Retaining Walls 7-2

7.3 State of Stress on a Soil Element Subjected to Stresses Induced by Wall Deformation 7-3

7.4 The relationship between Ka, Kp, and Ko 7-4

7.5 Development of Rankine Active and Passive Failure Zones for a Smooth Retaining Wall 7-6

7.7 Schematic Of Coulomb’s Theory Plane Failure Wedge of Soil 7-8

7.8 Comparison of Plane and Log-Spiral Failure Surfaces 7-10

7.9 Passive Coefficients for Sloping Wall with Wall Friction and Horizontal Backfill 7-11

7.10 Passive Coefficients for Vertical Wall with Wall Friction and Sloping Backfill 7-12

7.11 Lateral Pressure Coefficient Chart for Granular Soil with Sloping Backfill 7-13

7.12 General Distribution of Combined Active Earth Pressure and Water Pressure 7-14

7.13 Lateral Pressure Due to Surcharge Loadings (after USS Steel, 1975) 7-16

7.14 Potential Failure of a Rigid Retaining Wall 7-17

7.15 Design Criteria for Rigid Retaining Walls (NAVFAC 1986) 7-18

7.16 Typical Mode of Global Stability 7-20

7.17 Potential Source of Subsurface Water 7-22

7.19 Determining the Maximum and Minimum Pressures under the Base of the Cantilever Retaining Wall 7-23

7.20 Typical Failure Mode of a Flexible Wall 7-25

7.21 Type of Sheet Pile Walls 7-27

7.22 Lateral Pressures Distribution for Fixed-End Method of Design of Cantilever Sheet Pile Wall in Granular Soils 7-29

7.24 Various types of Anchoring for sheet pile walls 7-31


March 2009 7-1

7 RETAINING WALL

7.1 GENERAL Generally the main application of retaining wall is to hold back earth and maintain a difference in the elevation of the ground surface. The retaining wall is designed to withstand the forces exerted by the retained ground or “backfill” and other externally applied loads without excessive deformation or movement, and to transmit these forces safely to a foundation and to a portion of the restraining elements, if any, located beyond the failure surface. Figure 7.1 illustrated the forces acting on a retaining wall and some of the related terminology commonly used in retaining wall design. Special considerations are often necessary for retaining walls to be constructed close to land boundaries, particularly in urban areas. Land take requirement for construction often place limitations on the use of certain forms of earth retention. The cost of constructing a retaining wall is usually high compared with the cost of forming a new slope. Therefore, the need for a retaining wall should be assessed carefully during design.

Figure 7.1 Forces Acting On Retaining Wall And Common Terminology

7.2 TYPE OF RETAINING WALLS The rigidity or flexibility of a wall system is fundamental to the understanding of the development of earth pressures and the analysis of the wall stability. In simple terms, a wall is considered to be rigid if it moves as a unit in rigid body rotation and/or translation and does not experience bending deformation. Most gravity walls can be considered rigid walls. Flexible walls are those that undergo bending deformations in addition to rigid body motion. Such deformations result in a redistribution of lateral pressures from the more flexible to the stiffer portions of the system. Virtually all wall systems, except gravity walls, may be considered to be flexible.


7-2 March 2009

Some of the typical retaining walls are as shown in Figure 7.2

Figure 7.2 Type of Retaining Walls 7.3 SHEAR STRENGTH – LATERAL EARTH PRESSURE RELATIONSHIP The concept of lateral pressure is related to the effective stress and shear strength discussed in Chapter 3, Item 3.2 to 3.4. It is recommended that reader should review the principles of effective stress shear strength before proceeding further in this Chapter.

Counterfort wall

Sheet Piling

Sheet Pile Wall

Reinforced Soil Soil Nailing

Cantilever Gravity Element

Tied-back (Anchored) Braced


March 2009 7-3

The concept of lateral earth pressure acting on a wall can be explained based on the basic of the wall deformation. Consider an element of soil within a dry coarse-grained cohesionless soil mass. The geostatic effective stress on an element at any depth, z. would be as shown in Figure 7.3(a). Since the ground is not disturbed without any deformation, it is regarded as ‘at-rest’ condition. The coefficient of lateral pressure for this condition is termed as K0. Assume that a hypothetical, infinitely thin, infinitely rigid “wall” is inserted into the soil without changing the “at rest” stress condition in the soil as shown in Figure 7.3 (b). Now suppose that the hypothetical vertical wall move slightly to the left, i.e., away from the soil element as shown in Figure 7.3(c). In this condition, the vertical stress would remain unchanged. However, since the soil is cohesionless and cannot stand vertically on its own, it actively follows the wall. In this event, the horizontal stress decreases, which implies that the lateral earth pressure coefficient is less than Ko since the vertical stress remains unchanged. When this occurs the soil is said to be in the “active” state. The lateral earth pressure coefficient at this condition is called the “coefficient of active earth pressure”, Ka.

Figure 7.3 State of Stress on a Soil Element Subjected to Stresses Induced by Wall Deformation (a) In-situ vertical and horizontal stresses (b) Insertion of hypothetical infinitely thin and infinitely rigid

(c) Active contition of wall movement away from retained soil (d) Passive contition of wall movement toward retained soil

Now, instead of moving away from the soil, suppose the hypothetical vertical wall move to the right into the soil element as shown in Figure 7.3 (d). Again, the vertical stress would remain unchanged. However, the soil behind the wall passively resists the tendency for it to move, i.e., the horizontal stress would increase, which implies that the lateral earth pressure coefficient would become greater than Ko since the vertical stress remains unchanged. When this occurs the soil is said to be in the “passive” state. The lateral earth pressure coefficient at this condition is called the “coefficient of passive earth pressure,” Kp. The relationship between Ka, Kp, and Ko can best be illustrated graphically by Figure 7.4 below.

ph=Kopo

po po po po

ph=Kopo ph=Kapo ph=Kppo

δp δa


7-4 March 2009

Figure 7.4 The relationship between Ka, Kp, and Ko 7.4 LATERAL EARTH PRESSURE 7.4.1 At-Rest Lateral Earth Pressure The at-rest earth pressure condition in Figure 7.3(a) and (b) represents the lateral effective stress that exists in a natural soil in its undisturbed state. For cut walls constructed in near normally consolidated soils, the at-rest earth pressure coefficient, Ko, can be approximated by the equation (Jaky, 1944): Ko = 1 – sin φ′ (7.1)

where φ′ is the effective (drained) friction angle of the soil. The magnitude of the at-rest earth pressure coefficient is primarily a function of soil shear strength and degree of overconsolidation, which, as indicated in Chapter 4, may result from natural geologic processes for retained natural ground or from compaction effects for backfill soils. In overconsolidated soils, Ko can be estimated as (Schmidt, 1966):

Ko = (1 − sin ′)(OCR)

Ω (7.2)

where Ω is a dimensionless coefficient, which, for most soils, can be taken as sin φ′ (Mayne and Kulhawy, 1982) and OCR is the overconsolidation ratio. Typical values of K0 are as shown below: Normally consolidated clay, Ko = 0.55 to 0.65 Lightly overconsolidated clays (OCR ≤ 4) Ko = up to 1 Heavily overconsolidated clays (OCR > 4) Ko = > 2 (Brooker and Ireland, 1965) Sand Ko = 0.4 to 0.5

Kp (Passive limit)

Ko (at rest) (Not a failure limit)

Ka (Active limit) δ, lateral soil movement

K

δ

Kp ( Passive limit)

, lateral soil movement

Ko ( at rest )( not a failure limit )

Ka ( active limit )

K

δ

Kp ( Passive limit)



Ka ( active limit )

K

δ

Kp ( Passive limit)



Ka ( active limit )

K


March 2009 7-5

At-Rest condition may be appropriate for heavily preloaded, stiff wall systems. However, at-rest conditions are not typically used for flexible wall systems such as steel sheet-pile wall, where the wall undergoes some lateral deformation and designing to a requirement of zero movement is not practical. 7.4.2 Active and Passive Lateral Earth Pressures Active earth pressure (condition in Figure 7.3(c)) occurs when the wall moves away from the soil and the soil mass stretches horizontally sufficient to mobilize its shear strength fully, and a condition of plastic equilibrium is reached. The ratio of the horizontal component or active pressure to the vertical stress is the active pressure coefficient Ka. Passive earth pressure occurs when a soil mass is compressed horizontally, mobilizing its shear resistance fully. The ratio of the horizontal component of passive pressure to the vertical stress is the passive pressure coefficient, Kp. The amount of movement necessary to reach the plastic equilibrium conditions is dependent primarily on the type of backfill material. Some guidance on these movements is given in Table 7.1

Table 7.1 Wall Displacements Required to Develop Active and Passive Earth Pressures

Soil Type and Condition Necessary Displacement

Active Passive

Dense Cohesiveless 0.001H 0.02H

Loose Cohesiveless 0.004H 0.06H

Siff Cohesive 0.01H 0.02 H

Soft Cohesive 0.02H 0.04HNote : H = Wall Height

(Source: Wu, 1975)

There are two well-known classical lateral earth pressure theories i.e. Rankine’s and Coulomb’s. Each furnishes expressions for active and passive pressures for a soil mass at the state of failure. 7.4.2.1 Rankine’s Theory Rankine’s Theory is based on the assumptions that the wall introduces no changes in the shearing stresses at the surface of contact between the wall and the soil. It is also assumed that the ground surfaces is a straight line (horizontal or inclined straight line) and that a plane failure surface develops.

7-6

When theshown infollowing

Where

Ka

Kp

φ

Note tha

Ka

Kp

Thus, witassumed

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= the slopina = the activep = the passiv

φ = the effect

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thout consided to be triangn

Figure 7.

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March 20

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March 2009 7-7

pa = Ka p0= Ka γ z (7.7)

pa = Kp p0= Kp γ ζ (7.8) where p0 = Effective overburden pressure (unit length)= γh

pa = Active lateral earth pressures (unit length) pp = Passive lateral earth pressures (unit length) z = Depth below the ground surface h = Depth of tension crack (clayey soil only)

Figure 7.6 Triangular Lateral Force Distribution By Rankine Theory (a) For Granular Soil (b) For Cohesive Soil With Tension Crack Depth ‘H’ (Active Case)

For non- granular (c’ – φ ‘) soils, the lateral pressures are : Pa = Kaγz – 2c Ka (7.9) Pp = Kpγz + 2c Kp (7.10)

c = Cohesive strength of soil Theoretically, in soils with cohesion, the active earth pressure behind the wall becomes negative from the ground surface to a critical depth z where γh is less than 2c′ √ Ka. This critical depth is referred to as the “tension crack.” The active earth pressure acting against the wall within the depth of the tension crack is assumed to be zero. Unless positive drainage measures are provided, water infiltration into the tension crack may result in hydrostatic pressure on the retaining structure and should be full added to the lateral earth pressure.

2c tan (45°-Ø

2)

pa = rZ tan2(45-Ø

2)-2c tan(45°-Ø

2)

2c tan (45+Ø

2)

Pp = rZ tan2(45+Ø

2)+2c tan(45+Ø

2) Z Z

(b)

Z Z

pa=rZ tan2 (45-Ø

2)

Pp=rZ tan2 (45+Ø

2)

pp=rZKp

pa=rZKo

ß

ßKa = cosß

β β

β β

Ka = cosß β β

β β

Ka = 1Kp

H

Z Z

(a)

Z Z

H


7-8 March 2009

7.4.2.2 Coulomb’s Theory Coulomb Theory is also based on limit equilibrium of a plane wedge of soil. However, the theory takes into consideration the effects of wall friction, sloping wall face as well as the sloping backfill. The pressures calculated by using these coefficients are commonly known as the Coulomb earth pressures. Since Coulomb’s method is based on limit equilibrium of a wedge of soil, only the magnitude and direction of the earth pressure is found. Pressure distributions and the location of the resultant are assumed to be triangular. Coulomb’s coefficients of lateral pressures are as follows with their related terms and pressures diagrams shown in Figure 7.7

Ka = cos 2 - θ

cos 2θ cos θ+ δ sin - θ sin - βcos - δ cos - β

(7.11)

Kp = cos 2 + θ

cos 2θ cos θ - δ

(7.12)

Figure 7.7 Schematic Of Coulomb’s Theory Plane Failure Wedge of Soil (a) Active Condition (b) Passive Condition


March 2009 7-9

7.4.2.3 Effects of Wall Friction The magnitude and direction of the developed wall friction depends on the relative movement between the wall and the soil. In the active case, the maximum value of wall friction develops only when the soil wedge moves significantly downwards relative to the rear face of the wall. In some cases, wall friction cannot develop. These include cases where the wall moves down with the soil, such as a gravity wall on a yielding foundation or a sheet pile wall with inclined anchors, and cases where the failure surface forms away from the wall, such as in cantilever and counterfort walls. The maximum values of wall friction may be takes as follows : Timber, steel, precast concrete wall δ max. = Ø’/2 Cast in-situ concrete wall δ max. = 2 Ø’/3 Considerable structural movements may be necessary, however, to mobilize maximum wall friction, for which the soil in the passive zone needs to move upwards relative to the structure. Generally, maximum wall friction is only mobilized where the wall tends to move downwards, for example, if a wall is founded on compressible soil, or for sheet piled walls with inclined tensioned members. Some guidance on the proportion of maximum wall friction which may develop in various cases is given below (Teng)

δ = 200 concrete or brick walls = 150 uncoated sheetpile = 00 if wall tends to move downward together with the soil = 00 sheetpiling with small penertration or penetrated into soft or loose soil = 00 if backfill is subjected to vibratiion

In general, the effects of wall friction on Rankine and Coulomb methods of earth pressure computation are as follows:

a) The Rankine method cannot take account of wall friction. Accordingly, Ka is overestimated

slightly and Kp is under-estimated, thereby making the Rankine method conservative for most applications.

b) The Coulomb theory can take account of wall friction, but the results are unreliable for passive earth pressures for wall friction angle values greater than φ′/3 because the failure surface is assumed to be a plane. The failure wedges assumed in the Coulomb analysis take the form of straight lines as shown in Figure 7.8. However, this contrasted with the curved shapes of failure surface observed in many model tests. This assumption resulted in Ka being underestimated slightly and Kp being overestimated very significantly for large values of φ′.

In general, the effect of wall friction is to reduce active pressure. It is small and often disregarded. However, wall friction increases the value of Kp significantly and thus could yield lateral earth pressure that could be very large and could be unsafe as passive earth pressure forces are generally resisting forces in stability analysis

7-10

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March 20

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March 200

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RETAINING WA

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RETAINING WA

cal Wall with l, 1948; NAVF

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Wall Friction FAC, 1986b)

and Sloping

March 20

Backfill

009


March 2009 7-13

Figure 7.11 Lateral Pressure Coefficient Chart for Granular Soil with Sloping Backfill

7-14

7.4.3 In cases ground wis zero aequal todistributiis added any poinSince Kw

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March 20

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March 2009 7-15

∆ps

= K qs (7.13)

where ∆ps = increase in lateral earth pressure due to the vertical surcharge load

qs = vertical surcharge load applied at the ground surface, K = appropriate earth pressure coefficient.

When traffic is expected to come to within a distance from the wall face equivalent to one-half the wall height, the wall should be designed for a live load surcharge. The standard loadings for highway structures in are expressed in terms of HA and HB loading as defined in BS 5400 : Part 2 : 1978. In the absence of more exact calculations, the nominal load due to live load surcharge may be taken from Table 7.2.

Table 7.2 Suggested Surcharge Loads to be Used in the Design of Retaining Structures

Road class Type of live loading Equivalent surcharge

Urban trunk Rural trunk (Road likely to be regularly used by heavy industrial traffic)

HA + 45 units of HB 20kPa

Primary distributor Rural main road HA = 37 ½ units of HB 15kPa

District and local distributors Other rural roads Access Roads, Carparks

HA 10kPa

Footpaths, isolated from roads Play areas 5kPa

Note : 1. It is recommended that these surcharges be applied to the 1 in 10 year storm condition.

2. For footpaths not isolated from roadways, the surcharge applying for that road class should be used.

(Source: Public Works Department, 1977) Point loads, line loads, and strip loads are vertical surface loadings that are applied over limited areas as compared to surcharge loads. Hence, the increase in lateral earth pressure used for wall system design is not constant with depth as is the case for uniform surcharge loadings. These loadings are typically calculated by using equations based on elasticity theory for lateral stress distribution with depth and are as shown in Table 7.13. Lateral pressures resulting from these surcharges should be added explicitly to other lateral pressures.


7-16 March 2009

Figure 7.13 Lateral Pressure Due to Surcharge Loadings (after USS Steel, 1975)

March 200

7.5 Rigid retand the wretaining

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7-18 March 2009

Figure 7.15 Design Criteria for Rigid Retaining Walls (NAVFAC 1986)

Definitions B = width of the base of the footing tan δt = friction factor between soil and base W = weight at the baseof wall. Includes

weight of wall for gravity walls. Includes weight of the soil above footing for cantilever and counterfort walls

c = cohesion of the foundation soil ca = adhesion between concrete and soil δ = angle of wall friction Pp = passive resistance Location of Resultant, R Based on moments about toe (assuming Pp=0)

d = Wa+Pvg-Phb

W+Pv

Criteria for Eccentricity, e e = d-B

2 ; e≤B/6 for soils; e≤B/4 for rocks

Factors of Safety Against Sliding FSδ = W+Pv tanδb+caB

Ph≥1.5 min

Applied Stress at Base (qmax, qmin, qeq) qmax = W+Pv

B(1+ 6e

B)

qmin = W+Pv

B(1- 6e

B)

Equivalent uniform (Meyerhof) applied stress, qeq is given as follows: qeq = W+Pv

B' where B’ = B-2e

Use uniform stress, qeq, for soils and settlement analysis; use trapezoidal distribution with qmax and qmin for rocks and structural analysis Deep-seated (Global) Stability Evaluate global stability using guidance in Chap. 6 (Slope Stability)


March 2009 7-19

7.5.1 Sliding/Translational Stability The horizontal component of all lateral pressures tends to cause the wall to slide along the base of the wall (or along any horizontal section of a gravity and crib wall). If the passive resistance is neglected, the sliding force along the bottom of the wall is resisted by a horizontal force which consists of friction, adhesion or a combination of both. If the bottom of base slab is rough, as the case of concrete poured directly on soil, the coefficient of friction is equal to tan φ', (φ' is the angle of internal friction of the soil). Typical coefficients of friction are as follows:

Course-grained (without silt) 0.55 Course-grained (with silt) 0.45 Silt 0.35 Sound rock (with rough surface) 0.60

For cohesive soils the adhesion between the base slab and the soil is assumed to be equal to the cohesive strength of the clay and φ is assumed to be zero. The designer should consider the possibility of reduction in cohesive strength due to construction works such as excavation, exposure to surface water etc. If the retaining wall is supported on piles, the entire vertical and horizontal load should be assumed to be carried by piles. No frictional resistance and no adhesion should be assigned along the base slab. For checking the sliding factor of safety, the live load surcharge is usually not considered in the stabilising forces over the heel of the wall. Also, the passive resistance of the soil in front of the wall is commonly neglected in the stability analysis. If it is included in the computation, as in the case where the toe of wall is covered by a large depth of soil, its value should be reduced to take care of the high potential of the soil to be removed by erosion, future excavation, and tension cracks in cohesive soils. The minimum safety factor for sliding/translational stability shall be of minimum 1.5. The sliding stability can be increase by either increasing the overall weight of the retaining wall or providing sufficient passive lateral resistance of the wall. This can be done by introducing a wider base, construction of structural shear key and incorporating deep foundation support. 7.5.2 Overturning Stability The lateral pressure due to the backfill and surcharge tends to tip the retaining over about its toe. This overturning moment is stabilised by the weight of the wall and the weight of the soil above the base of the wall. The overturning stability of the wall is always the most critical potential mode of failure when the walls are underlain by weak soils. The minimum factor of safety against overturning is:

Fs = Sum of stabilizing moment Sum of overturning moment

≥2.0

To overcome the overturning stability, normally pile foundation is recommended. For some cases, ground improvement such as removal and replacement is adopted to increase the bearing capacity of the ground (provided the soft bearing ground is relatively thin). For passive resistance of the soil in front of the wall, designer should evaluate whether to ignore or to use a reduced value basing on the reason discussed in 7.5.1 above.


7-20 March 2009

7.5.3 Bearing Capacity Failure The computed vertical pressure at the base of the wall footing must be checked against the ultimate bearing capacity of the soil. The generalized distribution of the bearing pressure at the wall base is illustrated in Figure 7.15. Note that the bearing pressure at the toe is greater than that at the heel. The magnitude and distribution of these pressures are computed by using the applied loads shown in Figure 7.15. The equivalent uniform bearing pressure, qeq, should be used for evaluating the factor of safety against bearing capacity failure. The procedures for determining the allowable bearing capacity of the foundation soils can be found in Chapter 5 (Bearing Capacity) of this Volume. Generally, the factor of safety against bearing failure is defined as

Fs = qult

qeq ≥ 2.0

Where

q ult = ultimate bearing pressure q eq = equivalent uniform bearing pressure (as computed according to Figure 10.15)

7.5.4 Global Stability The overall stability shall be checked to avoid deep seated failure due to circular rotational or non-circular failure beyond the retaining wall. It must be checked with respect to the most critical failure surface. The minimum factor of safety for the overall stability shall be of minimum 1.5. A typical mode of circular rotational stability condition is illustrated in Figure 7.16 If global stability is found to be a problem, deep foundations or the use of lightweight backfill may be considered. Alternatively, measures can be taken to improve the shear strength of the weak soil stratum. Other wall types, such as an anchored soldier pile and lagging wall or tangent or secant pile wall, should also be considered in this case.

Figure 7.16 Typical Mode of Global Stability


March 2009 7-21

7.5.5 Selection of Backfill Materials The ideal backfill for a retaining is a free draining granular material of high shearing strength. However, the final choice of material should be based on the costs and availability of such materials balanced against the cost of more expensive walls. In general, the use of fine-grained clayey backfills is not recommended due to the following reasons:

a) Clays are subject to seasonal variations in moisture content and consequent swelling and shrinkage. This effect may lead to an increase in pressure against a wall when these soils are used as backfill.

b) As clays are subjected to consolidation, long terms settlement problems are considerably greater than with cohesionless materials.

c) For clay backfill, special attention must be paid to the provision of drainage to prevent the build-up of water pressure. Free draining cohesionless materials may not require the same amount of attention in this respect.

d) The wall deflection required to produce the active state in cohesive materials with a significant clay content may be up to 10 times greater than for cohesionless materials. This, together with the fact that the former generally have lower values of shearing strength, means that the amount of shear strength mobilized for any given wall movement is considerably lower for cohesive materials than for cohesionless materials. The corresponding earth pressure on the active side for a particular wall movement will therefore be higher if cohesive soil is used for backfill.

It is essential to specify and supervise the placing of backfill to ensure that its strength and unit weight properties agree with the design assumptions both for lateral earth pressure and dead weight calculations. In this regard, it is particularly important to ensure that the backfill behind a wall and on a slope is properly compacted. The backfill should normally be compacted in thin layers using light compaction plant so as not to minimize compaction loading on the wall. 7.5.6 Design Wall Drainage System Control of water is a key component of the design of earth retaining structures. Both subsurface water and surface water can cause damage during and/or after construction of the wall. Surface water runoff can destabilize a structure under construction by inundating the backfill. It can also destabilize a completed structure by erosion or by infiltrating into the backfill. Hence, adequate and proper design for surface water runoff is important to ensure the stability of the wall. Potential sources of subsurface water are surface water infiltration and groundwater as illustrated in Figure 7.17.


7-22 March 2009

Figure 7.17 Potential Source of Subsurface Water Drainage system design depends on wall type, backfill and/or retained soil type, and groundwater conditions. Drainage system components such as granular soils, prefabricated drainage elements and filters, are usually sized and selected based on local experience, site geometry, and estimated flows, although detailed design is only occasionally performed. Drainage systems may be omitted if the wall is designed to resist full water pressure. Drainage measures for fill wall systems and cut wall systems typically consist of the use of a free-draining material at the back face of the wall, with “weep holes” and/or longitudinal collector drains along the back face as shown in Figure 7.18. The collector drains may be perforated pipes or gravel drains. Where weepholes are used, BS 8002 specified that they should be at least 75 mm in diameter and at a spacing of not more than 1 m horizontally and 1 m to 2 m vertically.

Surface Water Infiltration

Drainage aggregate

Fill

Foundation Soil

Retained Fill

Groundwater

March 200

7.5.7 Determinshown in

09

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7-24 March 2009

The applicable soil properties and strength parameters are given as follows: Soil unit weight, γs = 17 kN/m3 Effective cohesion, c’ = 0 kN/m2 Effective angle of shearing resistance, φ’ = 40o Assume friction on the base of wall, δ = 30o Unit weight of concrete, γc = 23.5 kN/m3 And, water table is below base of wall. Solution: i. To determine the position of the base reaction, the moment of all forces about the heel of

the wall (X) are calculated as follows (Table 7.3 refers).

Table 7.3 Calculation Table

Force per m (kN) Arm (m) Moment per m (kNm)

(1) 0.22 x 40 x 5.40 = 47.5 2.70 128.2 (2) ½ x 0.22 x 17 x 5.402 = 54.6

Rh = 102.1

1.80 98.3

(Stem) 5.00 x 0.30 x 23.5 = 35.3 1.90 67.0 (Base) 3.00 x 0.40 x 23.5 = 28.2 1.50 42.3 (Soil) 5.00 x 1.75 x 17 = 148.8 0.875 130.2 (Load) 1.75 x 40 = 70.0

Rv = 282.3 0.875 61.3

M = 527.3 The active pressure is calculated on the vertical through the heel of the wall. No shear stresses act on this vertical, and therefore the Rankine theory (δ = 0) is used to calculate the active pressure using the pressure distribution as shown in Figure 1 above. Thus: For φ’ = 40 0 (and δ = 0), Ka = 0.22

Lever arm of base resultant, MRv

= 527.3282.3

= 1.81 i.e., the resultant acts within the middle third of the base. ii. Thus, eccentricity of base reaction, e = 1.81 – 1.50 = 0.31 m The maximum and minimum base pressures are given by:

Rv

B1± 6e

B

p = 282.3

31± 6x0.36

B3 = 94 (1 ± 0.72)

= 112 kN/m2 and 21 kN/m2

March 200

Thus the

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7-26 March 2009

7.6.2 Types of Flexible Walls The following retaining wall types are commonly used in Malaysia either to retain and/or support soils during excavations:

a) Sheet pile wall b) Soldier pile wall c) Contiguous bored pile / caisson wall d) Diaphragm wall

a) Sheet Pile Walls The sheet pile wall is used in many types of temporary and permanent structures. It is one of the most common methods used in the Department especially for the support and protection of river banks, water front construction, flood defence as well as temporary supports or containment for construction of hydraulic structures. Steel sheet piles are preferred mainly because of their ease of installation, length of service life and ability to be driven through water. However, they are not suitable when high bedrock or boulders prevent penetration to the required depth. When selecting sheet piles to be used, it is important to consider the drivability of the piles. The ability of the sheet pile to penetrate the ground depends on the section size of the pile and the type of the pile hammer used, as well as the ground conditions. It is difficult to drive sheet piles through soils with Standard Penetration Test (SPT) ‘N’ values greater than 50 (subjected to pile section). Further discussion on the basic principles in design of sheet pile wall are discussed in Item 7,6.3 below. b) Soldier Pile Wall Soldier pile wall has two basic components, soldier piles (vertical component) and lagging (horizontal component). Soldier piles provide intermittent vertical support and are installed before excavation commences. Due to their relative rigidity compared to the lagging, the piles provide the primary support to the retained soil as a result of the arching effect. Spacing of the piles is chosen to suit the arching ability of the soil and the proximity of any structures sensitive to settlement. A spacing of 2 – 3 m is commonly used in strong soils and no sensitive structures are present. The spacing is reduced to 1 – 2 m in weaker soils or near sensitive structures. c) Contiguous Bored Pile /Caisson Wall Replacement pile wall i.e., contiguous bored pile wall or caisson wall is the common excavation support system adopted in Malaysia. Generally, these types of wall are used as the permanent retaining wall system for basement construction and sometimes for high wall in hillside development. Bored piles or caisson piles are constructed continuously in a row to form retaining structures. A gap of approximately 75mm to 100mm is allowed between the piles. for ground with high ground water table or loose soils, grout columns are introduced between the gaps behind the wall system. For a better water tide conditions pressured grout columns can be used to minimize the water leakage. For caisson wall, it is commonly used at areas with limited working space; where big machinery i.e., boring rig and excavator are not possible.

March 200

d) D Diaphragthe soil rpanels toDiaphraginstallatiosurround 7.6.3 7.6.3.1 The sheeform of s

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7-28

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March 20

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March 2009 7-29

b) Fixed-end Method A wall designed using Fixed-end principles is embedded sufficiently deep enough so that at the foot of the wall, both translation and rotation are prevented and fixity is assumed. This is the condition assumed in the design of a cantilever sheet pile wall. Figure 7.22 (a) and (b) illustrated the deflected shape of a cantilever sheet pile together with the conventional and simplified pressure distributions used for design. An example on the application of this method in Cantilever sheet pile wall desiGn is given in Item 7.6.3.3 below.

A tie or prop may also be provided at the upper part of the wall as shown in Figure 7.23 (a), (b) and (c). The effect of toe fixity is to create a fixed end moment in the wall, reducing the maximum bending moment for a given set of conditions but at the expense of increased pile length. The design method used (whether Free-end or Fixed-end Method) should also consider the effects of hydrostatic pressures and surcharge loads, which are usually added to that due to the soils.

Figure 7.23 Fixed-end Methpod of Design of Prop Sheet Pile Wall in ranular soils (a) Deflected shape of wall (b) Idealized lateral preswsure distribution (c) Simplified Lateral Pressure Distribution

Dredge Line Deflected shape of pile

(a) (b)

Deflected shape of pile

(a) (b) (c)

Figure 7.22 Lateral pressures distribution for Fixed-end Method of design of cantilever sheet pile wall in granular soils: (a) Idealized distribution (b) Simplified distribution


7-30 March 2009

7.6.3.3 Design of Anchor - General In the analysis of anchored steel sheet pile wall, whether using the Fixed-end or Free-end method, the tie or strut force, F , per unit length of the wall can be obtained. The restaining anchor must be designed to take the required force, F. In general, the types of anchor used in sheet pile wall are:

a) Anchor plates and beams (deadman) Figure b) Tie backs c) Vertical anchor piles d) Anchor beam supported by batter (compression or tension) piles

These anchors are as shown in Figure 7.24 (a), (b), (c), and (d) respectively.

(a) Anchor plates and beams

(b) Tie backs (c) Vertical anchor piles


March 2009 7-31

Figure 7.24 Various types of Anchoring for sheet pile walls (a) Anchor Plate or Beams; (b) Tie Back; (c) Vertical Anchor Pile; (d) Anchor Beam with Batter Piles

The above figures also illustrated the proper locations for placement of various types of anchors. Readers can refer to ‘Principles of Getechnical Engineering’ by M. B. Das for further guidance on the design of the various types of anchors. 7.6.3.4 Some Considerations on Sheet Pile Wall Design a) Selection of Analysis Method

Designers must be careful when selecting the design approach to adopt i.e., the Fixed-end or Free end method. Walls installed in soft cohesive soils, may not generate sufficient pressure to achieve fixity and in those soils it isrecommended that free earth conditions are assumed. Fixed earth conditions may be appropriate where the embedment depth of the wall is taken deeper than that required to satisfy lateral stability, i.e. to provide an effective groundwater cut-off or adequate vertical load bearing capacity. However, where driving to the required depth may be problematic, assumption of free earth support conditions will minimise the length of pile to be driven and ensure that the theoretical bending moment is not reduced by the assumption of fixity. When designing a wall involving a significant retained height and multiple levels of support, the overall pile length will often be sufficient to allow the designer to adopt fixed earth conditions for the early excavation stages and take advantage of reduced bending moment requirements. b) Construction Sequence

The design of tied-back or braced system should also consider the sheet pile design requirements at each and every stages of the construction sequence, i.e. excavation, strutting, anchoring and lowering of ground water table. This construction sequence shall be detailed in the construction drawings as wrong construction sequence may cause large changes in the bending moment, shear stress and overall stability of the wall. c) Permissible Stress of Steel Sheet Pile

In the design of temporary sheet pile wall, the permissible steel stresses for the structural design of the sheet pile can be increased slightly. For instance, Piling Handbook, Archelor Group suggested that the permissible steel stresses for temporary works (wall to last not more than 3 months) shown in Table 7.3 be used in the structural design in the sheet piles and other steel components of the wall such as walins, struts and tie rod.

(d) Anchor beam supported by batter (compression or tension) piles


7-32 March 2009

Table 7.4 Permissible Steel Stress of Sheet Pile

Class of Work Steel grade to EN10248

S270GP (N/mm2)

S355GP (N/mm2)

Permanent 180 230 Temporary 200 260

d) Design of Cofferdam

Cofferdam is a retaining structure, usually temporary in nature, which is used to temporary support the sides of deep excavation such as in the construction of multi-level basements and trenches for construction of bridge abutment, piers and instalation of deep pipe culverts. Its method of construction involved instalation of vertical steel sheet piles to required depth and as excavation works progress, a system of wales and struts or prestressed tiebacks (anchors) is installed. The earth lateral pressures for the multi-level cofferdam cannot be calculated by the classical pressures theories ( Rankine, Coulomb and wedge theories). Readers are advised to refer to literatures such as Foundation Design by W.C. Teng or Steel Sheet Piling Design Manual, USS for design of this type of wall. In addition, the effects of seepage forces and piping need to be considered especially where high differential water levels existing between the inner and outer face of the wall. Seepage forces and piping or boiling effects can lead to wall instability by reducing passive earth pressure, and in more severe cases, can cause liquifaction or ‘quick sand' condition. BS8004 1981 provides some guides on the minimum depth of cut-off for cohesionless soils (Table 9, pg 47)and shown belows:

Width, W Depth of cut-off, D 2Y or more Y 0.5Y

0.4Y 0.5Y 0.7Y

Notes: a) The stability of the wall could

be increased by increasing the seepage flow path.

b) A narrow trench needs a deeper cut-off.

c) Value of D obtained to be compared with value for stability.

Table 9 ( BS8004 )Idea is to increase seepage flow path.

Note that a narrow trench needs a deeper cut-off.

Value of D obtained to be compared with value for stability.

W

Y

D

GWL

Table 9 ( BS8004 )Idea is to increase seepage flow path.

Note that a narrow trench needs a deeper cut-off.

Value of D obtained to be compared with value for stability.

W

Y

D

GWL


March 2009 7-33

e) Engineering Software

Many commercial softwares are also available to facilitate the analysis of retaining wall. Most of these software are capable of analyzing more complex and complicated situation e.g. basement excavation where high accuracy is required. Some computer programs used the numerical solutions to model the soil-structure interaction analysis. Some of these softwares include WALLAP by Geosolve, ReWaRD by Geocentrix, FREW by OASYS and many others are available. Finite element software such as PLAXIS, SIGMA/W are also becoming increasing more popular as they are able to simulate the response of the wall and the soils under various design loadings and construction sequence.

7.6.3.3 Cantilever Steel Sheet Pile Retaining Wall - Example

A wall is to be built to support a retained height of 3.2m of sandy soils. The effective wall height = 3.2m + 10% = 3.52m say 3.5m (unplanned excavation allowance is 10% with 0.5m maximum). Minimum surcharge loading = 10 kN/m2. Based on Carquot & Kerisel Chart for Ka and Kp (Fig. 7.9) Loose fine sand Ka = 0.3 Kp = 0.746 x 6.5 (Ø = 30°, δ/Ø = -0.5, Reduction Factor for Kp = 0.746 – From Fig. 7.9)

Compact fine sand Ka = 0.26 Kp = 0.7 x 8.3 = 5.8 (Reduction Factor for Kp = 0.70)

SURCHARGE 10 kN/m2

Loose Fine Sand γ = 17.5 kN/m3 γsat = 19.1 kN/m3

= 30°

Compact Fine Sand γ = 18.5 kN/m3 γsat = 19.81 kN/m3

= 33° γw = 9.81 kN/m3

GWL GWL

4.50

m

6.0

m

1.0

m

0.30

m U

npla

nned

3.

2 m

-δ/Ø = -0.5 for both soil layer

TYPICAL SECTION

Overburden kN/m2 Active Passive

Water Soil Water Soil 0.00

0.00

0.00

58.86

10.00

88.75

107.25

167.25

0.00

17.50

36.00

96.00

0.00

0.00

0.00

58.86

GWL


7-34 March 2009

Note: As ground water levels are the same on both active and passive sides of the wall, pressures due to water are ignored. Active pressures Pa at 0.00 m below G.L. in loose sand = 0.3 x 10.00 = 3.0 kN/m2 Pa at 4.50 m below G.L. in loose sand = 0.3 x 88.75 = 26.63 kN/m2 Pa at 4.50 m below G.L. in loose sand = 0.260 x 88.75 = 27.89 kN/m2 Pa at 5.50 m below G.L. in loose sand = 0.260 x 167.25 = 43.49 kN/m2 Pa at 11.50 m below G.L. in loose sand = 0.260 x 167.25 = 43.49 kN/m2 Passive pressures Pp at 3.50 m below G.L. in loose sand = 4.8 x 0.00 = 0.00 kN/m2 Pp at 4.50 m below G.L. in loose sand = 4.8 x 17.50 + 0.00 = 84.00 kN/m2 Pp at 4.50 m below G.L. in loose sand = 5.8 x 17.50 + 0.00 = 101.50 kN/m2 Pp at 5.50 m below G.L. in loose sand = 5.8 x 36.00 = 208.80 kN/m2 Pp at 11.50 m below G.L. in loose sand = 5.8 x 96.00 = 556.80 kN/m2

March 200

09

Chapter 7 RRETAINING WAALL

7-3

35


7-36 March 2009

Take moments about the toe at 7.022m depth Active force

Force (kN/m)

Moment about toe (kNm/m)

3.0 x 6 = 18.00 x 3.0 = 54.00 23.63 x 4.5 x 1/2 = 53.17 x 3.00 = 159.50 20.07 x 1.000 = 20.07 x 1.00 = 20.07 4.81 x 1.000 x ½ = 2.41 x 0.833 = 2.01 24.88 x 0.32 = 7.96 x 0.16 = 1.27 0.83 x 0.32 x ½ = 0.133 x 0.11 = 0.014 101.74 236.86 Passive force

Force (kN/m)

Moment about toe (kNm/m)

84.0 x 1000 x ½ = 42 x 1.65 = 69.30 101.50 x 1000 = 101.50 x 1.0 = 101.50 106.80 x 1.000 x ½ = 53.40 x 0.833 = 44.48 208.30 x 0.50 = 104.15 x 0.167 = 17.36 29.0 x 0.5 x ½ = 7.25 x 0.167 = 1.21 308.30 233.85 Since the passive moment is marginally less than the active moments length is OK. To correct the error caused by the use of the simplified method in the depth below the point of equal active and passive pressure is increased by 20% to give the pile penetration. Let the point of equal pressure be (3.5 + d) below ground level

Then 84

1.00 x d = 3.0 +

23.634.5

x (3.5 + d)

Therefore d = 18.38

84 – 5.25 = 0.233m

Hence the required pile length = 3.50 + 0.233 + 1.2 x (2.50 – 0.233) = 6.45m say 6.50m. Zero shear occurs at 4.77m below ground level (where the area of the active pressure diagram above the level equals the area of the passive pressure diagram above the level). Take the moments about and above the level of zero shear (point O): kNm/m 3.0 x 4.77 x ½ x 2.385 = 17.06 23.63 x 4.5 x ½ x 1.77 = 94.11 0.056 x 0.27 x ½ x 0.009 = 0.00 20.08 x 0.27 x 0.135 = 6.73 -84.00 x 1.000 x ½ x 0.6 = -25.20 -101.50 x 0.27 x 0.091 = -2.49 -28.84 x 0.27 x ½ x 0.09 = -0.35 83.86


March 2009 7-37

Maximum bending moment = 83.86 kNm/m. A partial factor of 1.2 is applied to give the ultimate load. Section modulus of pile required = 1.2 x 83.36 x 103 / 270 = 373 cm3/m Hence use PU6 piles (z=600 cm3/m) not less than 6.50m long in S270GP. However the designer will need to check the sustainability of the section for driving and durability.


7-38 March 2009

REFERENCES [1] Bishop A.V and Henkel D.J., The Measurement of Soil Properties in the Triaxial Test, E.Arnold, 1962. [2] Bowles, J.E. Foundation Analysis and Design. (Fourth edition). McGraw-Hill International, New York, 1992, 1004 p. [3] Brown, R.W., (1996) Practical foundation Engineering Handbooks, Mcgraw-Hill [4] BSI. Eurocode 7: Geotechnical Design – Part 1: General Rules (BS EN 1997-1 : 2004). British Standards Institution, London, 2004, 117 p. [5] Carter M. & Symons, M.V., Site Investigations and foundations Explained, Pentech Press, London [6] CGS, “Canadian Foundation Engineering Manual”, (Third edition). Canadian Geotechnical Society, Ottawa, 1992, 512 p. [7] Das, B.M., Principles of Geotechnical Engineering, PWK-Kent Publishing Company , Boston,MA., 1990 [8] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C., NAVFAC DM-7.1, May 1982, Soil Mechanics [9] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C.,NAVFAC DM-7.2, May 1982, Foundations and Earth Structures [10] DID Malaysia, Geotechnical Guidelines for D.I.D. works [11] DID Malaysia, Retaining Wall [12] GCO (1990) Review of Design Method for Excavation, Geotechnical Control Office, Hong Kong [13] GEO (1993). Guide to Retaining Wall Design (Geoguide 1). (Second edition). Geotechnical Engineering Office, Hong Kong, 217 p. [14] Harry R.Cedergreen, Seepage, Drainage and Flownet, John Wiley nd Sons. [15] Holtz, R.D., Kovacs, W.D. An Introduction to Geotechnical Engineering, Prentice-Hall, Inc. New Jersey [16] Ladd C.C., Foott R., Ishihara K., Schlosser F., and Roulos H.G., "Stress Deformation and Strength Characteristics", State of the Art Report, Session I, IX ICSMFE, Tokyo, Vol. 2, 1971, pp. 421 - 494. [17] Lambe T.W. and Whitman R.V., "Soil Mechanics", John Wiley 8: Sons, 1969 [18] McCarthy D.J., "Essentials of Soil Mechanics and Foundations". [19] Nayak N. V. I II Foundation Design Manual. Dhanpat Rai a Sons I 1982.


March 2009 7-39

[20] Peck R.B Hanson W.E. and Thornburn R.H., “Foundation Engineering", John Wiley and Sons, 1974. [21] Poulos, H.G., Carter, J.P. & Small, J.C. (2002). Foundations and retaining structures – research and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. [22] Research and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. [23] Smith C.N., "Soil Mechanics for Civil and Mining Engineers". [24] Teng W.C., "Foundation Design", Prentice Hall, 1984. [25] Terzaghi, K. & Peck, R.B. (1967). Soil Mechanics in Engineering Practice. (Second edition). Wiley, New York, 729 p. [26] United Bureau States Department of the Interior, "Design of Small Dams” Bureau of Reclamation, Oxford and IBH Publishing Co., 1974. [27] Vesic, A.S. (1975). Bearing capacity of shallow foundations. Foundation Engineering Handbook, edited by Winterkorn, H.F. & Fang, H.Y., Van Nostrand Reinhold, New York, pp 121-147.


7-40 March 2009


CHAPTER 8 GROUND IMPROVEMENT


March 2009 8-i

Table of Contents


List of Tables ......................................................................................................................... 8-ii


8.1 INTRODUCTION .......................................................................................................... 8-1

8.2 SOIL IMPROVEMENT TECHNIQUES ............................................................................... 8-2

8.2.1 Removal and Replacement .............................................................................. 8-2

8.2.2 Surcharging ................................................................................................... 8-3

8.2.3 SUB SURFACE DRAINAGE IMPROVEMENT SYSTEM ........................................... 8-3

8.2.3.1 Vertical Drainage System ................................................................. 8-4

8.2.3.2 Sand Drain System .......................................................................... 8-5

8.2.3.3 Prefabricated Vertical Drain (PVD) .................................................... 8-5

8.2.4 Vibro-Floatation ............................................................................................. 8-6

8.2.4.1 Vibro Compaction ............................................................................ 8-6

8.2.4.2 Vibro Replacement (Stone Column)................................................... 8-7

8.2.5 DEEP SOIL MIXING (LIME COLUMN) ................................................................ 8-8

8.2.5.1 Mix Design ...................................................................................... 8-9

8.2.6 Dynamic Compaction ...................................................................................... 8-9

8.2.7 Some Additional Considerations ...................................................................... 8-10

REFERENCES ....................................................................................................................... 8-12

APPENDIX 8A: DESIGN OF VERTICAL DRAINAGE SYSTEM ....................................................... 8A-1


8-ii March 2009

List of Tables


8.1 Typical Properties and Test Standards Specified For Vertical Drain 8-6


8.1 Distribution of Alluvium Deposits In Peninsular Malaysia 8-1

8.2 Typical Drainage Directions in Soft Soil During Consolidation Process 8-4

8.3 Typical Drainage Direction with Vertical Drainage System in Soft Soil during Consolidation Process 8-4

8.4 Typical Schematic Diagram For Vertical Sand Drain System In Embankment Construction on Soft Ground 8-5

8.5 Prefabricated Vertical Drain 8-5

8.6 Relationships between Particle Size and Available Vibro Techniques 8-6

8.7 The Schematic Process of Vibro Compaction 8-7

8.8 Schematic Showing the Installation of Stone Columns (Dry Method) 8-8

8.9 Mixer Paddle Used In Deep Soil Mixing 8-9

8.10 Dynamic Compaction 8-10

8.11 Relationships between U and Tv 8A-2

8.12 Relationship Of Uh and Tv For Horizontal/Radial Drainage 8A-2

8.13 Relationship of F(n) and D/dw 8A-4

8.14 Design Chart for Horizontal Consolidation 8A-5

March 2

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2009

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8-2 March 2009

Fortunately, engineers and contractors have developed methods of coping with these problematic soils and have successfully built many large structures on very poor sites. Among the methods used (either individually or in combination) include:-

a) Support the structures on deep foundations that penetrate through the weak soils b) Support the structure on shallow foundations and design them to accommodate the weak

soils c) Use a floating foundation, either deep or shallow d) Remove the poor material and replace with good materials. This approach is only effective

if the poor soil material is relatively thin and good replacement soil materials can be easily found on site.

e) Improve the engineering properties of the soils. Various methods of ground improvement techniques are available which basically aim to reduce the pore water pressure, reduce the volume of voids in the soil, add stronger materials and additives (such as lime or cementitious grout) to enhance its soil properties

f) Avoid the poor ground either by re-alignment or shifting the location of the structures (if availability of land is not a constraint)

The main objectives of ground improvements are to:-

• Reduce settlement of structures • Improve shear strength and bearing capacity of shallow foundations • Increase factor of safety against possible slope failure of embankments and dams. • Reduce shrinkage and swelling of soils

The most common techniques often used in our country for solving and stabilizing soft ground problems are listed below:- a) Structure support system using the shallow foundation or deep foundation and incorporating

either partially or fully floating foundation principle. Readers are advised to refer to Chapter 5 and Chapter 9 for shallow foundation and deep foundation respectively.

b) Soil improvement and stabilization works include

i) Removal and replacement ii) Surcharging iii) Sub-surface drainage improvement system iv) Vibro floatation v) Deep mixing – Lime column vi) Dynamic compaction

8.2 SOIL IMPROVEMENT TECHNIQUES 8.2.1 Removal and Replacement Sometimes poor soils can simply be removed and replaced with good quality compacted fill. This alternative is especially attractive if the thickness of the deposit is small, the groundwater table is deep and good quality fill material is readily available. If the soil is inorganic and not too wet, then it probably is not necessary to haul it away. Such soils can be improved by simply compacting them. In this case, the contractor excavates the soil until firm ground is exposed and then places the excavated soil back in its original location, compacting it in lifts. This technique is often called removed and re-compaction. If necessary, the soil can be reinforced with geosynthetics to spreads the applied load over a larger area, thus reducing the change in effective stress and reducing the consolidation settlement as well as increasing the bearing capacity.


March 2009 8-3

Removal and Replacement (or re-compaction) technique is one of the most common and relatively less expensive methods used in infrastructures development such as road and earthworks construction. However, its usage is limited or constraint by:-

a. Thickness of unsuitable soft soil Often, this technique is only applicable to soft soil layers with thickness less than 3 meter.

Thick removal may require massive temporary shoring to be in place and end up being more costly.

b. Availability of replacement material Availability of replacement material is an important factor as it will govern the overall

construction cost. Sometimes, light weight material such as Expanded Polystyrene System (EPS) is used as an alternative replacement material to minimize excessive consolidation settlement and bearing failure of thick fill area.

8.2.2 Surcharging Covering poor soils with a temporary surcharge fill, as shown in Figure 8.3, causes them to consolidate more rapidly. When the temporary fill is removed, some or all of the soil is now overconsolidated, and thus stronger and less compressible. Often, preloading (by surcharging) has been used to improve saturated silts and clays because these soils are most conducive to consolidation under static loads. Sandy and gravelly soils respond better to vibratory loads. If the soil is saturated, the time required for it to consolidate depends on the ability of the excess pore water to move out of the soil voids (see the discussion of consolidation theory in Chapter 4). This depends on the thickness of the soil deposit, its coefficient of permeability, and other factors, and can be estimated using the principles of soil mechanics. The time required could range from only a few weeks to thirty years or more. Allowable construction period is an important factor to determine the height of surcharge. Lesser surcharge height will require longer surcharge time. For condition where high embankment or surcharge load is required, stage construction can be introduced to avoid bearing failure during construction. Consolidation process during stage construction will increase soil strength in order to allow higher load at the next stages. The consolidation process can be accelerated by an order of magnitude or more by installing vertical drains in the natural soil, as discussed in Item 8.2.3. These drains provide a pathway for the excess water to escape more easily. Preloading is less expensive than some other soil improvement techniques, especially when the surcharge soils can be moved from place to place, thus preloading the site in sections. Vertical drains, if needed will increase the cost substantially. 8.2.3 Sub Surface Drainage Improvement System In general sub-drainage system, either horizontal or vertical (or both), can be used to accelerate consolidation process by reducing drainage path. These drainage systems provide a pathway for the excess water to escape more easily. Vertical drainage system is the most commonly used system for embankment constructed on soft soil (provided there are no sand layers or lenses exist in the ground) and the directional flows of these drains are as shown in Figure 8.2. The length of the drainage path is determined by the thickness of the soft soil or by the existence of any drainage layers such as sand layers or lenses. The longer the drainage path, the longer the time required to achieve the desired degree of consolidation.


8-4 March 2009

Figure 8.2 Typical Drainage Directions in Soft Soil During Consolidation Process 8.2.3.1 Vertical Drainage System The introduction of a grid of vertical drains will reduce the traveling distance of the water path during consolidation process (refer Figure 8.3), thus increases the rate of consolidation. The presence of any natural permeable layers or lenses will further enhance and facilitates horizontal water flow toward the vertical drains. This minimizes the excess water pressure generated during and after construction and increases the rate of settlement. Generally there are 2 common vertical drainage systems available in the market, namely:-

a) Sand drain system b) Prefabricated vertical drain (PVD) system

Figure 8.3 Typical Drainage Direction with Vertical Drainage System In Soft Soil During Consolidation Process

March 2

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March 2009

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March 2009 8-9

binder. The binder is injected into the soil in a dry form. The moisture in the soil is utilized for the binding process, resulting in an improved soil with higher shear strength and lower compressibility. The removal of the moisture from the soil also results in an improvement in the soft soil surrounding the mixed soil.

Figure 8.9 Mixer Paddle Used In Deep Soil Mixing

Typical applications of the deep soil mixing method include foundations of embankment fill for highway and railway, slope stabilization, stabilization of deep excavation and foundations for housing development. The anticipated amounts of binding agents commonly used are approximately 100 – 150 kg/m3 in silty clay and clayey silt materials. The strength develops differently over time depending on the type of soil, amount of binder and proportion used. In most cases, the strength starts to increase after a few hours and then continues to increase rapidly during the first week. In normal cases, approximately 90% of the final strength is reached after about three weeks. 8.2.5.1 Mix Design Detailed site investigation and laboratory tests are required to determine the optimum lime content for soil stabilization. In general, lime stabilization is suitable for ground with low sulphide and organic content. It is also effective for silty ground with low plasticity. The optimum lime percentage is approximately 3% but increases with water content. However if lime content exceeded the optimum content, shear strength of treated ground will be reduced. The increase in the shear strength after improvement varies, and ranges from 5-10 kPa to 100kPa. Generally shear strength increment reduces with increment of liquid limit. The soil strength increase gradually through the pozzolonic reaction between lime, aluminate and silicate in the soil (clay). The percentage of clay shall be more than 20%. For normal case, the mixture of silt and clay shall be greater than 35% and plasticity shall be greater than 10%. If the percentage of clay does not fulfill the condition above, cement and fly ash shall be added. For soil improvement using lime mixing in organic soil, shear strength increment is rather small. Usually, gypsum is added to unslaked lime to stabilize the organic soil. The mixture is of approximately ¼ to ½ of gypsum to ¾ ~ ½ unslaked lime. 8.2.6 Dynamic Compaction Dynamic compaction consists of using a heavy tamper that is repeatedly raised and dropped with a single cable from varyingn heights to impact the ground. The mass of the tampers generally ranges from 20 tonnes to 200 tonnes and drop height range from 20 to 40m. The energy is generally applied in phases on a grid pattern over the entire area using single or multiple passes. Following

8-10

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PROVEMENT

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March 2009

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March 2009 8-11

c) Soils subjected to improvement works are usually very soft in nature. Standard Penetration Test

(SPT) is not suitable for soft soil layer. It is advisable to retrieve undisturbed soil samples from the ground for laboratory tests which include Undrained Unconsolidated (UU) Triaxial test and One Dimensional Consolidation Test using Odeometer. In addition, in-situ tests such as Vane Shear test and Piezocone are recommended in soft soils sensitive to disturbance such as marine clay is highly recommended.

d) Transition zone shall be provided in the ground improvement design if the project used more than one type of ground improvement methods. This is most crucial if the ground improvement methods pose a different allowable long term settlement, e.g., bridge and bridge approach, culverts etc.

e) Due to the complexities and uncertainties of the ground conditions as well as the simplification of design formulae in the analysis and design, it is strongly recommended that the instrumentation monitoring scheme shall be provided during the construction works for design verification purposes. Some provisions in the Bill of Quantities shall also be provided to cater for any design changes during construction.


8-12 March 2009

REFERENCES [1] ASCE (1987). Soil Improvement – A ten Year Update, Geotechnical Special Publication No. 12, edited by J.P. Welsh. [2] Bowles, J.E. (1988). Foundation Analysis and Design, 4th ed., McGraw-Hill, New York. [3] Broms, B.B. (1993). Lime Stabilization. “Chapter 4 in Ground Improvement, edited by M.P. Moseley, CRC Press, Boca Raton, Florida, pp. 65-99. [4] Broms, B.B., and Forssblad, L. (1969). “Vibratory Compaction of Cohesionless Soils. “Proceedings of the Seventh International Conference on Soil Mechanics and Foundation Engineering, Specialty Session No. 2, pp. 101-118. [5] Broomhead, D., and Jasperse, B.H. (1992). “Shallow Soil Mixing- a Case History. “Grouting, Soil Improvement and Geosynthetic, edited by R.H. Borden, R.D. Holtz, and I. Juran, ASCE Geotechnical Special Publication no. 3o, vol. 1, pp. 564 – 576. [6] Brown, R.W., (1996) Practical foundation Engineering Handbooks, Mcgraw-Hill [7] Coduto, D. P., (2001) Foundation Design – Principles and Practices, Prentice Hill Inc. [8] Das, B.M. (1983). Advanced Soil Mechanics, Hemisphere Publishing, New York. [9] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C.,NAVFAC DM.-7.3, April 1983, Soil Dynamics, Deep Stabilization and Special Geotechnical Construction [10] Duncan, J.M. & Poulos, H.G. (1981). Modern techniques for the analysis of eng [11] ineering problems in soft clay. Soft Clay Engineering, Elsevier, New York, pp 317-414. [12] EM 1110-2-1913. Design and Construction of Levees, U.S. Army Corp of Engineer, Washington, DC. [13] FHWA (1979). Soil Stabilization in Pavement Structures- a User’s Manual, Report no. FHWA-IP-80-2, Federal Highway Administration, Washington, D.C., October. [14] Hausmann, M.R. (1990). Engineering Principles of Ground Modification, McGraw-Hill, New York. [15] Koerner R.M . Construction and Geotechnical Method in Foundation Engineering, McGraw Hill, 1985. [16] McCarthy D.J., Essentials of Soil Mechanics and Foundations. [17] Mesri G., discussion of New Design Procedure for stability of Soft Clays. by Charles C. Ladd and Roger Foott, Journal of the Geotechnical Engineering Division, ASCE, Vol.101, No. GT4. Froc. Paper 10664. April 1975. pp. 409 - 412. [18] Nayak N. V. I II, Foundation Design Manual. Dhanpat Rai a Sons I 1982. [19] Peck R.B Hanson W.E. and Thornburn R.H., Foundation Engineering, John Wiley and Sons, 1974.


March 2009 8-13

[20] O.G., and Metcalf, J.B. (1973), Soil Stabilization: Principles and Practice, John Wiley & Sons, New Ingles. [21] PCA(1979). Soil-Cement Construction handbook, Portland Cement Association, Skokie, Illinois. [22] Sherwood, P.T.(1962). Effect of Sulfates on Cement-and Lime-Stabilized Soils. Highway Research Board Buletin No. 353: Stabilization of Soils with Portland Cement, Washington, D.C., pp. 98-107. Also in Roads and Road Construction, vol. 40, February, pp. 34-40. [23] Sokolovich, V.E., and Semkin, V.V. (1984), Chemical Stabilization of Loess Soils. Soil Mechanics and Foundation Engineering, vol. 21, no. 4, July-August, pp. 8-11. [24] Teng W.C., Foundation Design, Prentice Hall, 1984. [25] Terzaghi, K. & Peck, R.B. (1967). Soil Mechanics in Engineering Practice. (Second edition). Wiley, New York, 729 p. [26] Thomson, M.R. (1966). Shear Strength and Elastic Properties of Lime-Soil Mixtures. Highway Research Record No. 139: Behaviour Characteristics of Lime-Soil Mixtures, highway Research Board, Washington, D.C., pp. 1-14. [27] Thonson, M.R. (1969). Engineering Properties of Soil-Mistures. Journal of Materials, ASTM, vol. 4, no. 4, December. [28] TRB (1987). Lime Stabilization: Reactions, Properties, Design, and Construction, State of the Art Report 5, Transportation Research Board, Washington, D.C.


8-14 March 2009



March 2009 8A-1

APPENDIX 8A: DESIGN OF VERTICAL DRAINAGE SYSTEM The principal objective of soil pre consolidation, with or without PVD, is to achieve a desired degree of consolidation within a specified period of time. The design of pre consolidation with PVDs requires the evaluation of drain and soil properties (both separately and as a system) as well as the effects of installation. For one dimensional consolidation with drains, only consolidation due to one dimensional (vertical) seepage to natural drainage boundaries is considered. The degree of consolidation can be measured by the ration of the settlement at any time to the total primary settlement that will (or is expected to) occur. This ratio is referred to as Ū, the average degree of consolidation. By definition, one dimensional consolidation is considered to result from vertical drainage only, but consolidation theory can be applied to horizontal or radial drainage as well. Depending on the boundary conditions consolidation may occur due to concurrent vertical and horizontal drainage. The average degree of consolidation, Ū, can be calculated from the vertical, horizontal or combined drainage depending on the situation considered. With Vertical drains the overall average degree of consolidation, Ū, is the result of the combined effects of the horizontal (radial) and vertical drainage. The combined effect is given by:-

Ū = 1 – ( 1 – Ūh) (1 – Ūv) (8.1) where, Ū = overall average degree of consolidation Ūh = average degree of consolidation due to horizontal (radial) Drainage Ūv = average degree of consolidation due to vertical drainage. The graph of Ū vs log time for both the vertical and horizontal drainage in shown in Figure 8.11 and Figure 8.12 respectively.

8A-2

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PROVEMENT

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March 2009

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March 2009 8A-3

t = (D2/8Ch) F(n) ln (1/(1- Ūh)) (8.2)

where, t = time required to achieve Ūh Ū = average degree of consolidation due to horizontal drainage. D = diameter of the cylinder of influence of the drain (drain influence zone) Ch = coefficient of consolidation for horizontal drainage F(n) = Drain spacing factor = ln (D/d) – ¾ D = diameter of a circular drain Equation 8.2 was further modified by Hasbo to be applied to band-shape PVD and to include consideration of disturbance and drain resistance effects.

t = (D2/8Ch) (F(n) + Fs + Fr) ln (1/(1- Ūh)) (8.3) where, t = time required to achieve Ūh Ū = average degree of consolidation at depth z du to horizontal drainage D = diameter of the cylinder of influence of the drain (drain influence zone) Ch = coefficient of consolidation for horizontal drainage

F(n) = Drain spacing factor = ln (D/dw) – ¾ D = diameter of a circular drain dw = equivalent diameter Fs = factor for soil disturbance = ((kh/ks ) – 1) ln (ds/dw)

kh = the coefficient of permeability in the horizontal direction in the undisturbed soil ks = the coefficient of permeability in the horizontal direction in the disturbed soil ds = diameter of the idealized disturbed zone around the drain Fr = factor for drain resistance = πz (l – z) (kh/qw) z = distance below top surface of the compressible soil later L = effective drain length; length of drain when drainage occurs at one end only; half

length of drain when drainage occurs at both ends qw = discharge capacity of the drain (at gradient = 1.0)

Equation 8.3 can be simplified to the ideal case by ignoring the effect of soil disturbance and drain resistance (Fs and Fr = 0) the resulting ideal case equation is equivalent to Barron’s solution: t = (D2/8Ch) F(n) ln (1/(1- Ūh)) (8.4) Therefore, in the ideal case, the time for a specified degree of consolidation simplifies to be a function of soil properties (Ch), design requirement (Ūh) and design variables (D, dw).


8A-4 March 2009

Figure 8.13 Relationship of F(n) and D/dw Figure 8.13 shows the relationship of F(n) to D/dw for the ideal case. Within a typical range of D/dw, F(n) ranges from approximately 2 to 3. The theory of consolidation with radial drainage assumes that the soil is drained by a vertical drain with circular section. The radial consolidation equations include the drain diameter, d. A band shape PVD drain must therefore be assigned as “equivalent diameter”, dw. For design purposes, it is reasonable to calculate the equivalent diameter as:- dw = (2(a+b)/π) (8.5) where,

a = width of the band – shaped drain cross section b = thickness of a band-shaped drain cross section

Equation A8.5 can be further simplified to dw = (a + b) /2 (8.6)

March 2

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PROVEMENT

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8A-6 March 2009

Table 8.1 Typical Properties and Test Standards Specified For Vertical Drain

Criteria Properties Standard

General Thickness ASTM D5199 Constructability Tensile Strength (dry and Wet)

Grab Strip Wide Width

ASTM D4132 ASTM D1182 ASTM D5035

Tear Strength ASTM D4533 Puncture resistance ASTM D4833 Abrasion resistance ASTM D4881 Ultra violet stability ASTM D4355

Hydraulic Permeability / permittivity ASTM D4491 Apparent opening size (O95) ASTM D4751 Discharge capacity ASTM D4711

CHAPTER 9 FOUNDATION ENGINEERING


March 2009 9-i

Table of Contents


List of Tables ...................................................................................................................... 9-iii

List of Figures ..................................................................................................................... 9-iii

9.1 INTRODUCTION .......................................................................................................... 9-1

9.2 DEEP FOUNDATION ..................................................................................................... 9-2

9.2.1 General ......................................................................................................... 9-2

9.2.2 Classification of Piles ....................................................................................... 9-2

9.2.2.1 Precast Reinforced Concrete Piles ....................................................... 9-2

9.2.3 Pile Foundation Design .................................................................................... 9-6

9.2.3.1 General ............................................................................................ 9-6

9.2.3.2 Design Philosophies ........................................................................... 9-6

9.2.3.4 Pile Capacity ..................................................................................... 9-8

9.2.4 Pile Loading Tests ........................................................................................ 9-13

9.2.4.1 General .......................................................................................... 9-13

9.2.4.2 Timing of Pile Tests ......................................................................... 9-14

9.2.4.3 Static Pile Loading Tests .................................................................. 9-14

9.2.5 Equipment ................................................................................................... 9-17

9.2.5.1 Measurement of Load ...................................................................... 9-17

9.2.5.2 Measurement of Pile Head Movement ............................................... 9-19

9.2.5.3 Test Procedures .............................................................................. 9-21

9.2.5.4 Instrumentation .............................................................................. 9-24

9.2.5.5 Interpretation of Test Results ........................................................... 9-25

9.2.6 Dynamic Loading Tests ................................................................................. 9-27

9.2.6.1 General .......................................................................................... 9-27

9.2.6.2 Test Methods .................................................................................. 9-27

9.2.6.3 Methods of Interpretation ................................................................ 9-28

9.2.6.4 Recommendations on the Use of Dynamic Loading Tests .................... 9-29

9.3 LATERALLY LOADED PILES ......................................................................................... 9-29

9.3.1 Introduction ................................................................................................. 9-29

9.3.2 Lateral Load Capacity of Pile .......................................................................... 9-31

9.3.3 Inclined Loads .............................................................................................. 9-39

9.3.4 Raking Piles in Soil ........................................................................................ 9-39

9.3.5 Lateral Loading ............................................................................................ 9-40


9-ii March 2009

9.3.5.1 General .......................................................................................... 9-40

9.3.5.2 Equivalent Cantilever Method ........................................................... 9-41

9.3.5.3 Subgrade Reaction Method .............................................................. 9-41

9.3.5.4 Elastic Continuum Method ................................................................ 9-43

9.4 PILE GROUP .............................................................................................................. 9-45

9.4.1 General ....................................................................................................... 9-45

9.4.2 Minimum Spacing of Piles ............................................................................. 9-46

9.4.3 Ultimate Capacity of Pile Groups .................................................................... 9-46

REFERENCES ..................................................................................................................... 9-48


March 2009 9-iii

List of Tables


9.1 Advantages and Disadvantages of Machine-dug Piles 9-4

9.2 Advantages and Disadvantages of Hand-dug Caissons 9-5

9.6 Tolerance of Installed Piles 9-46

List of Figures


9.1 Types of Foundation 9-1

9.2 Estimation of Negative Skin Friction by Effective Stress Method 9-13

9.3 Typical Arrangement of a Compression Test using Kentledge 9-15

9.4 Typical Arrangement of a Compression Test using Tension Piles 9-16

9.6 Typical Instrumentation Scheme for a Vertical Pile Loading Test 9-21

9.7 Typical Load Settlement Curves for Pile Loading Tests (Tomlinson, 1994) 9-26

9.8 Failure Modes of Vertical Piles under Lateral Loads (Broms, 1914a) 9-30

9.9 Coefficients Kqz and Kcz at Depth z for Short Piles Subject to Lateral Load (Brinch Hansen, 1911) 9-33

9.10 Ultimate Lateral Resistance of Short Piles in Granular Soils (Broms, 1914a) 9-34

9.11 Ultimate Lateral Resistance of Long Piles in Granular Soils (Broms, 1914b) 9-35

9.12 Influence Coefficients for Piles with Applied Lateral Load and Moment (Flexible Cap or Hinged End Conditions) (Matlock & Reese, 1910) 9-37

9.13 Influence Coefficients for Piles with Applied Lateral Load (Fixed against Rotation at Ground Surface) (Matlock & Reese, 1910) 9-38

9.14 Analysis of Behaviour of a Laterally Loaded Pile Using the Elastic Continuum Method (Randolph, 1981a) 9-44


9-iv March 2009


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9-2 March 2009

9.2 DEEP FOUNDATION 9.2.1 General Deep foundation is usually used when tructural load is relatively high and/or the ground condition does not allow for shallow foundation system. Sometimes due to high load, required spread footing are too large and not economical. For some special structures, i.e., bridge pier, dock etc, pile foundation is adopted because the foundation is subjected to scour or undermining. Generally deep foundation system is also preferable where the structures are subjected to high uplift force or lateral force.

9.2.2 Classification of Piles There are many types of pile classification adopted. In general, piles can be classified according to:- a) The type of material forming the piles, b) The mode of load transfer, c) The degree of ground displacement during pile installation and d) The method of installation. Pile classification in accordance with material type (e.g. steel and concrete) has drawbacks because composite piles are available. A classification system based on the mode of load transfer will be difficult to set up because the proportion of shaft resistance and end-bearing resistance that occurs in practice usually cannot be reliably predicted. In the installation of piles, either displacement or replacement of the ground will predominate. A classification system based on the degree of ground displacement during pile installation, such as that recommended in BS 8004 (BSI, 1981) encompasses all types of piles and reflects the fundamental effect of pile construction on the ground which in turn will have a pronounced influence on pile performance. Such a classification system is therefore considered to be the most appropriate. In this document, piles are classified into the following four types: (a) Large-displacement piles, which include all solid piles, including precast concrete piles, and steel

or concrete tubes closed at the lower end by a driving shoe or a plug, i.e. cast-in-place piles, large diameter spun pile etc.

(b) Small-displacement piles, which include rolled steel sections such as H-piles and open-ended tubular piles. However, these piles will effectively become large-displacement piles if a soil plug forms.

(c) Replacement piles, which are formed by machine boring, grabbing or hand-digging. The excavation may need to be supported by bentonite slurry, or lined with a casing that is either left in place or extracted during concreting for re-use.

(d) Special piles, which are particular pile types or variants of existing pile types introduced from time to time to improve efficiency or overcome problems related to special ground conditions.

9.2.2.1 Precast Reinforced Concrete Piles Precast reinforced concrete piles are common nowadays in Malaysia. These piles are commonly in square sections ranging from about 250 mm to about 450 mm with a standard length varies from 1m to 12m. The lengths of pile sections are often dictated by the practical considerations including transportability, handling problems in sites of restricted area and facilities of the casting yard In general, and the maximum allowable axial loads is subjected to the structural capacity designed by the manufacturer and it can be up to about 1 000kN. These piles can be lengthened by coupling together during installation. Joining method commonly adopted in Malaysia is using wielding of the end plate of the piles.


March 2009 9-3

This type of pile is not suitable for driving into ground that contains a significant amount of boulders or corestones and very hard sand lenses. i) Precast Prestressed Spun Piles Precast prestressed spun concrete piles used in Malaysia are closed-ended tubular sections of 400 mm to 1000 mm diameter with maximum allowable axial loads up to about 3000 kN. Special large diameter spun piles with diameter greater than 1000mm are also available but the demand is low. Pile sections are normally 12 m long and are usually welded together using steel end plates. Precast prestressed spun concrete piles require high-strength concrete and careful tight QA/QC control during manufacture. Casting is usually carried out in a factory where the curing conditions can be strictly regulated. Special manufacturing processes such as compaction by spinning or autoclave curing can be adopted to produce high strength concrete up to about 75 MPa. Such piles may be handled more easily than precast reinforced concrete piles without damage. Steam curing is usually adopted in the casting yard to shorten casting time and to ensure the quality of the pile. ii) Small-Displacement Piles Small-displacement piles are either solid (e.g. steel H-piles) or hollow (open-ended tubular piles, i.e., GI pipes) with a relatively low cross-sectional area. This type of pile is usually installed by percussion method. However, a soil plug may be formed during driving, particularly with tubular piles, and periodic drilling out may be necessary to reduce the driving resistance. A soil plug can create a greater driving resistance than a closed end, because of damping on the inner-side of the pile. Bakau pile is considered to be a small displacement pile. However, due to the conservation of the mangrove forest and the coastal line of Malaysia. Bakau piles are not allowed to be used special permit is required if imported bakau pile is used. iii) Replacement Piles Replacement or bored piles are mostly formed by machine excavation. When constructed in condition with high ground water table, the pile bore will need to be supported using steel casings, concrete rings or drilling fluids such as bentonite slurry, polymer mud, etc to avoid collapsing of drilled hole. Excavation of the pile bore may also be carried out by hand-digging in the dry; and the technique developed in Hong Kong involving manual excavation is known locally as hand-dug caissons. Machine-dug piles are formed by rotary boring, or percussive methods of boring, and subsequently filling the hole with concrete. Piles with 100 mm or less in diameter are commonly known as small-diameter piles. Piles greater than 1000 mm diameter are referred to as large-diameter piles. a) Machine Bored Piles The advantages and disadvantages of machine-dug piles are summarized in Table 9.1.


9-4 March 2009

Table 9.1 Advantages and Disadvantages of Machine-dug Piles

Advantages Disadvantages

i. No risk of ground heave induced by pile driving.

ii. Length can be readily varied. ii. Spoil can be inspected and compared with

site investigation data. v. Structural capacity is not dependent on

handling or driving conditions. v. Can be installed with less noise and vibration

compared to displacement piles. vi. Can be installed to great depths. vii. Can readily overcome underground

obstructions at depths.

a. Risk of loosening of sandy or gravelly soils during pile excavation, reducing bearing capacity and causing ground loss and hence settlement.

b. Susceptible to bulging or necking during concreting in unstable ground.

c. Quality of concrete cannot be inspected after completion except by coring.

d. Unset concrete may be damaged by significant water flow.

e. Excavated material requires disposal, the cost of which will be high if it is contaminated.

f. Base cleanliness may be difficult to achieve, reducing end-bearing resistance of the piles.

b) Mini / Micro Bored Piles Mini-piles generally have a diameter between 100 mm and 400 mm. One or more high yield steel bars are provided in the piles. In Malaysia, used high yield steel pipes are commonly used as the reinforcement for micro piles. Construction can be carried out typically to about 10 m depth or more, although verticality control will become more difficult at greater depths. Mini-piles are usually formed by drilling rigs with the use of down-the-hole hammers or rotary percussive drills. They can be used for sites with difficult access or limited headroom and for underpinning. In general, they can overcome large or numerous obstructions in the ground. Mini-piles are usually embedded in rock sockets. Given the small-diameter and high slenderness ratio of mini-piles, the load is resisted largely by shaft resistance. The lengths of the rock sockets are normally designed to match the pile capacity as limited by the permissible stress of steel bars. A mini-pile usually has four 50 mm diameter high yield steel bars and has a load-carrying capacity of about 1375 kN. Where mini-piles are installed in soil, the working load is usually less than 700 kN but can be in excess of 1 000 kN if post grouting is undertaken using tube-a-manchette. Pile cap may be designed to resist horizontal loads. Alternatively, mini-piles can be installed at an inclination to resist the horizontal loads. c) Large Diameter Bored Piles Large-diameter bored piles are used in Malaysia to support heavy column loads of tall buildings and highways structures such as viaducts. Typical sizes of these piles range from 1 m to 3 m, with lengths up to about 80 m and working loads up to about 45,000 kN. The working load can be increased by socketing the piles into rock or providing a bell-out at pile base. The pile bore is supported by temporary steel casings or drilling fluid, such as bentonite slurry.


March 2009 9-5

d) Hand Dug Caissons Hand-dug caissons are not very common in Malaysia. For the past two decades, it has been widely used in project with limited working space and for hillside development. Their diameters typically range from 1.2 m to 2.5 m, with an allowable load of up to about 25000 kN. The advantages and disadvantages of hand-dug caissons are summarised in Table 9.2. Hand-dug caisson shafts are excavated using hand tools in stages with depths of up to about 1 m, depending on the competence of the ground. Dewatering is facilitated by pumping from sumps on the excavation floor or from deep wells. Advance grouting may be carried out to provide support in potentially unstable ground. Each stage of excavation is lined with in-situ concrete rings (minimum 75 mm thick) using tapered steel forms which provide a key to the previously constructed rings. When the diameter is large, the rings may be suitably reinforced against stresses arising from eccentricity and non-uniformity in hoop compression. Near the bottom of the pile, the shaft may be belled out to enhance the load-carrying capacity. Examples of situations where the use of caissons should be avoided include: • Coastal reclamation sites with high groundwater table, • Sites underlain by cavernous marble, • Deep foundation works (e.g. In excess of say 50 m), • Landfill or chemically-contaminated sites, • Sites with a history of deep-seated ground movement, • Sites in close proximity to water or sewerage tunnels, • Sites in close proximity to shallow foundations, and • Sites with loose fill having depths in excess of say 10 m.

Examples of situations where hand-dug caissons may be considered include:

• Steeply-sloping sites with hand-dug caissons of less than 25 m in depth in soil, and • Sites with difficult access or insufficient working room where it maybe impracticable or unsafe

to use mechanical plant.

Table 9.2 Advantages and Disadvantages of Hand-dug Caissons

Advantages Disadvantages a) As (a) to (e) for machine-dug piles. b) Base materials can be inspected. c) Versatile construction method requiring

minimal site preparation and access. d) Removal of obstructions or boulders is

relatively easy through the use of pneumatic drills or, in some cases, explosives.

e) Generally conducive to simultaneous excavation by different gangs of workers.

f) Not susceptible to programme delay arising from machine down time.

g) Can be constructed to large-diameters.

a) As (a), (c) and (e) for machine-dug piles. b) Hazardous working conditions for workers

and the construction method has a poor safety record.

c) Liable to base heave or piping during excavation, particularly where the groundwater table is high.

d) Possible adverse effects of dewatering on adjoining land and structures.

e) Health hazards to workers, as reflected by a high incidence rate of pneumoconiosis and damage to hearing of caisson workers.


9-6 March 2009

9.2.3 Pile Foundation Design 9.2.3.1 General Methods based on engineering principles of varying degrees of sophistication are available as a framework for pile design. All design procedures can be broadly divided into four categories: (a) Empirical 'rules-of-thumb', (b) Semi-empirical correlations with in-situ test results, (c) Rational methods based on simplified soil mechanics or rock mechanics theories, and (d) Advanced analytical (or numerical) techniques. A judgment has to be made on the choice of an appropriate design method for a given project. In principle, in choosing an appropriate design approach, relevant factors that should be considered include: (a) The ground conditions, (b) Nature of the project, and (c) Comparable past experience. 9.2.3.2 Design Philosophies The design of piles should comply with the following requirements throughout their service life: • There should be adequate safety against failure of the ground. The required factor of safety

depends on the importance of the structure, consequence of failure, reliability and adequacy of information on ground conditions, sensitivity of the structure, nature of the loading, local experience, design methodologies, number of representative preliminary pile loading tests.

• There should be adequate margin against excessive pile movements, which would impair the serviceability of the structure.

a) Global Factor of Safety Approach

The conventional global factor of safety approach is based on the use of a lumped factor applied notionally to either the ultimate strength or the applied load. This is deemed to cater for all the uncertainties inherent in the design.


March 2009 9-7

The conventional approach of applying a global safety factor provides for variations in loads and material strengths from their estimated values, inaccuracies in behavioural predictions, unforeseen changes to the structure from that analysed, unrecognised loads and ground conditions, errors in design and construction, and acceptable deformations in service. b) Limit State Design Approach A limit state is usually defined as 'any limiting condition beyond which the structure ceases to fulfil its intended function'. Limit state design considers the performance of a structure, or structural elements, at each limit state. Typical limit states are strength, serviceability, stability, fatigue, durability and fire. Different factors are applied to loads and material strengths to account for their different uncertainty. c) Recommended Factors Of Safety The following considerations should be taken into account in the selection of the appropriate factors of safety:

(i) There should be an adequate safety factor against failure of structural members in accordance with appropriate structural codes.

(ii) There must be an adequate global safety factor on ultimate bearing capacity of the ground. Terzaghi et al (1991) proposed the minimum acceptable factor of safety to be between 2 and 3 for compression loading. The factor of safety should be selected with regard to importance of structure, consequence of failure, the nature and variability of the ground, reliability of the calculation method and design parameters, extent of previous experience and number of loading tests on preliminary piles. The factors as summarised in Table 9.3 for piles in soils should be applied to the sum of the shaft and end-bearing resistance (HONG KONG GEO 2001).

(iii) The assessment of working load should additionally be checked for minimum 'mobilisation' factors fs and fb on the shaft resistance and end-bearing resistance respectively as given in Table 9.5.

(iv) Settlement considerations, particularly for sensitive structures, may govern the allowable loads on piles and the global safety factor and/or 'mobilisation' factors may need to be higher than those given in (ii) & (iii) above.

(v) Where significant cyclic, vibratory or impact loads are envisaged or the properties of the ground are expected to deteriorate significantly with time, the minimum global factor of safety to be adopted may need to be higher than those in (ii), (iii) and (iv) above.

(vi) Where piles are designed to provide resistance to uplift force, a factor of safety should be applied to the estimated ultimate pile uplift resistance and should not be less than the values given in Table 9.4.


9-8 March 2009

Table 9.3 Minimum Global Factors of Safety for Piles in Soil and Rock

Material Mobilization Factor for Shaft Resistance, fs

Mobilization Factor for End-bearing Resistance, fb

Granular Soils Clays

1.5 1.2

3 – 5 3 – 5

Notes: 1. Mobilization factors for end-bearing resistance depend very much on construction. Recommended minimum factors assume good workmanship without presences of debris giving rise to a ‘soft’ toe and are based on available local instrumented loading tests on friction piles in granitic saprolites. Mobilization factors for end-bearing resistance. The higher the ratio, the lower is the mobilization factor.

2. Noting that the movements required to mobilize the ultimate end-bearing resistance are about 2% to 5% of the pile diameter for driven piles and about 10% to 20% of the pile diameter for bored piles, lower mobilization factor may be used for driven piles.

3. In stiff clays, it is common to limit the peak average shaft resistance to 100 kPa and the mobilized base pressure at working load to a nominal value of 550 to 600 kPa for settlement considerations, unless higher values can be justified by loading tests.

4. Where the designer judges that significant mobilization of end-bearing resistance cannot be relied on at working load due to possible effects of construction, a design approach which is sometimes advocated (e.g. Toh et al, 1989, Brooms & Chang, 1990) is to ignore the end-bearing resistance altogether in determining the design working load with a suitable mobilization factor on shaft resistance alone (e.g. 1.5). .End-bearing resistance is treated as an added safety margin against ultimate failure and considered in checking for the factor of safety against ultimate failure.

5. Lower mobilization factor for end-bearing resistance may be adopted for end-bearing piles provided that it can be justified by settlement analyses that the design limiting settlement can be satisfied.

9.2.3.4 Pile Capacity a) Design of Geotechnical Capacity in soil

Pile capacity can be divided into 2 main components, namely;

• Shaft resistance; Qs • End bearing resistance; Qb

The ultimate capacity of the pile is the sum of both the shaft resistance and the end bearing resistance; Qult = Qs + Qb (9.6) As for allowable pile capacity; Qallow = Qs/Fs + Qb/Fb (9.7) Where,

Fs = safety factor for shaft resistance. The common Fs adopted in design is 2.0 Fb = safety factor for end bearing. The common Fb ranges from 2.0 to 3.0 subjected to

availability and sufficiency of soil parameters. Higher safety factor shall be used when limited soil information is made available. As for bored pile, normally Qb is ignored.


March 2009 9-9

The design of pile geotechnical capacity commonly used can be divided into two major categories namely:

i) Semi-empirical Method ii) Simplified Soil Mechanics Method

i) Semi-Empirical Method Piles are constructed in tropical soils that generally have complex soil characteristics. The current theoretically based formulae do not consider the effects of soil disturbance, stress relief and partial reestablishment of ground stresses that occur during the installation of piles; therefore, the sophistication involved in using such formulae may not be necessary. Semi-empirical correlations have been extensively developed relating both shaft resistance and base resistance of piles to N-values from Standard Penetration Tests (SPT ’N’ values). In the correlations established, the SPT ’N’ values generally refer to uncorrected values before pile installation. The commonly used correlations for bored piles are as follows: fs = Ks x SPT ’N’ (in kPa) (9.8) fb = Kb x SPT ’N’ (in kPa) (9.9) Where:

Ks = Ultimate shaft resistance factor Kb = Ultimate base resistance factor

SPT’N’ = Standard Penetration Tests blow counts (blows/300mm) Toh et al. (1989) reported that the average Ks obtained varies from 5 at SPT ’N’ 20 to as low as 1.5 at SPT ’N’=220. Chang & Broms (1991) suggests that Ks of 2 for bored piles in residual soils of Singapore with SPT ’N’<150. For base resistance, Kb values available varies significantly indicating difficulty in obtaining proper and consistent base cleaning during construction, especially for bored piles. It is very dangerous if the base resistance is relied upon when the proper cleaning of the base cannot be assured. From back-analyses of test piles, Chang & Broms (1991) shows that Kb equals to 30 to 45 and Toh et al. (1989) reports that Kb falls between 27 and 10 as obtained from the two piles that were tested to failure. ii) Simplified Soil Mechanics Methods Generally the simplified soil mechanics methods for bored pile design can be classified into fine grained soils (e.g. clays, silts) and coarse grained soils (e.g. sands and gravels).

Fine Grained Soils The ultimate shaft resistance (fs) of bored piles in fine grained soils can be estimated based on the semi-empirical undrained method as follows: fs = α x su (9.10)


9-10 March 2009

Where :

α = adhesion factor su = undrained shear strength (kPa)

Whitaker & Cooke (1911) reports that the α value lies in the range of 0.3 to 0.1 for stiff overconsolidated clays, while Tomlinson (1994) and Reese & O’Neill (1988) report α values in the range of 0.4 to 0.9. The α values for residual soils of Malaysia are also within this range. Where soft clay is encountered, a preliminary value of 0.8 to 1.0 is usually adopted together with the corrected undrained shear strength from the vane shear test. This method is useful if the bored piles are to be constructed on soft clay near river or at coastal area. The value of ultimate shaft resistance can also be estimated from the following expression: fs = Kse x σv ’ x tan φ’ (9.11) Where :

Kse = Effective Stress Shaft Resistance Factor = [can be assumed as Ko] σv ’ = Vertical Effective Stress (kPa) φ’ = Effective Angle of Friction (degree) of fined grained soils.

However, this method is not popular in Malaysia and limited case histories of back-analysed Kse values are available for practical usage of the design engineer.

Although the theoretical ultimate base resistance for pile in fine grained soil can be related to undrained shear strength as follows; fb = Nc x su (9.12) Where:

Nc = bearing capacity factor

Note: it is not recommended to include base resistance in the calculation of the bored pile geotechnical capacity due to difficulty and uncertainty in base cleaning.

Coarse Grained Soils The ultimate shaft resistance (fs) of piles in coarse grained soils can be expressed in terms of effective stresses as follows: fsu = β x σv’ (9.13) Where:

β = shaft resistance factor for coarse grained soils. The β values can be obtained from back-analyses of pile load tests. The typical β values of piles in loose sand and dense sand are 0.15 to 0.3 and 0.25 to 0.1 respectively based on Davies & Chan (1981).


March 2009 9-11

c) Negative Skin Friction Piles installed through compressible materials (e.g. fill or marine clay) can experience negative skin friction. This occurs on the part of the shaft along which the downward movement of the surrounding soil exceeds the settlement of the pile. Negative skin friction could result from consolidation of a soft deposit caused by dewatering or the placement of fill. The dissipation of excess pore water pressure arising from pile driving in soft clay can also result in consolidation of the clay. The magnitude of negative skin friction that can be transferred to a pile depends on (Bjerrum, 1973): (a) Pile material, (b) Method of pile construction, (c) Nature of soil, and (d) Amount and rate of relative movement between the soil and the pile

In determining the amount of negative skin friction, it would be necessary to estimate the position of the neutral plane, i.e. the level where the settlement of the pile equals the settlement of the surrounding ground. For end-bearing piles, the neutral plane will be located close to the base of the compressible stratum.

Calculation of Negative Skin Friction Design of negative skin friction should include checks on the structural and geotechnical capacity of the pile, as well as the downward movement of the pile due to the negative skin friction dragging the pile shaft (CGS, 1992; Fellenius, 1998). A pile will settle excessively when geotechnical failure occurs. As the relative displacement between the soil and the pile shaft is reversed, the effect of negative skin friction on pile shaft would be eliminated. Therefore, the geotechnical capacity of the pile could be based on the shaft resistance developed along the entire length of pile. The drag load need not be deducted from the assessed geotechnical capacity when deciding the allowable load carrying capacity of the pile. On the other hand, the structural capacity of the pile should be sufficient to sustain the maximum applied load and the drag load. The drag load should be computed for a depth starting from the ground surface to the neutral plane. The estimation of downward movement of the pile (i.e. downdrag) requires the prediction of the neutral plane and the soil settlement profile. At the neutral plane, the pile and the ground settle by the same amount. The neutral plane is also where the sustained load on the pile head plus the dragload is in equilibrium with the positive shaft resistance plus the toe resistance of the pile. The total pile settlement can therefore be computed by summing the ground settlement at the neutral plane and the compression of the pile above the neutral plane (Figure 9.2). For piles founded on a relatively rigid base (e.g. on rock) where pile settlement is limited, the problem of negative skin friction is more of the concern on the structural capacity of the pile. This design approach is also recommended in the Code of Practice for Foundations (BD, 2004a) for estimating the effect of negative skin friction. For friction piles, various methods of estimating the position of the neutral plane, by determining the point of intersection of pile axial displacement and the settlement profile of the surrounding soil, have been suggested by a number of authors (e.g. Fellenius, 1984). However, the axial displacement at the pile base is generally difficult to predict without pile loading tests in which the base and shaft responses have been measured separately. The neutral plane may be taken to be the pile base for an end-bearing pile that has been installed through a thick layer of soft clay down


9-12 March 2009

to rock or to a stratum with high bearing capacity. The method includes the effect of soil- structure interaction in estimating the neutral plane and drag load on a pile shaft. Alternatively, the neutral plane can be conservatively taken as at the base of the lowest compressible layer (BD, 2004a). The mobilised negative skin friction, being dependent on the horizontal stresses in the ground, will be affected by the type of pile. For steel H-piles, it is important to check the potential negative skin friction with respect to both the total surface area and the circumscribed area relative to the available resistance (Broms, 1979). The effective stress or β method may be used to estimate the magnitude of negative skin friction on single piles (Bjerrum et al, 1919; Burland & Starke, 1994). In general, it is only necessary to take into account negative skin friction in combination with dead loads and sustained live load, without consideration of transient live load or superimposed load. Transient live loads will usually be carried by positive shaft resistance, since a very small displacement is enough to change the direction of the shaft resistance from negative to positive, and the elastic compression of the piles alone is normally sufficient. In the event where the transient live loads are larger than twice the negative skin friction, the critical load condition will be given by (dead load + sustained live load + transient live load). The above recommendations are based on consideration of the mechanics of load transfer down a pile (Broms, 1979) and the research findings (Bjerrum et al, 1919; Fellenius, 1972) that very small relative movement will be required to build up and relieve negative skin friction, and elastic compression of piles associated with the transient live load will usually be sufficient to relieve the negative skin friction. Caution needs to be exercised however in the case of short stubby piles founded on rock where the elastic compression may be insufficient to fully relieve the negative skin friction. In general, the customary local assumption of designing for the load combination of (dead load + full live load + negative skin friction) is on the conservative side. Poulos (1990b) demonstrated how pile settlement can be determined using elastic theory with due allowance for yielding condition at the pile/soil interface. If the ground settlement profile is known with reasonable certainty, due allowance may be made for the portion of the pile shaft over which the relative movement is insufficient to fully mobilise the negative skin friction (i.e. movement less than 0.5% to 1% of pile diameter).

March 200

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9-14 March 2009

There are two main types of pile loading tests, namely static and dynamic loading tests. Static loading tests are generally preferred because they have been traditionally used and also because they are perceived to replicate the long-term sustained load conditions. Dynamic loading tests are usually carried out as a supplement to static loading tests and are generally less costly when compared with static loading tests. The failure mechanism in a dynamic loading test may be different from that in a static loading test. The Statnamic loading test is a quasi-static loading test with limited local experience. In this test, a pressure chamber and a reaction mass is placed on top of the pile. Solid fuel is injected and burned in the chamber to generate an upward force on the reaction mass. An equal and opposite force pushes the pile downward. The pile load increases to a maximum and is then reduced when exhausted gases are vented from the pressure chamber. Pile displacement and induced force are automatically recorded by laser sensors and a load cell. The load duration for a Statnamic loading test is relatively long when compared with other high energy dynamic loading tests. While the additional soil dynamic resistance is usually minimal and a conventional static load-settlement curve can be produced, allowance will be required in some soil types such as soft clays. 9.2.4.2 Timing of Pile Tests For cast-in-place piles, the timing of a loading test is dictated by the strength of the concrete or grout in the pile. Weltman (1980b) recommended that at the time of testing, the concrete or grout should be a minimum of seven (7) days old and have strength of at least twice the maximum applied stress. With driven piles, there may be a build-up of pore water pressure after driving. Lam et al (1994) reported that for piles driven into weathered meta-siltstone the excess pore water pressure built up during driving took only one and a half days to dissipate completely. Results of dynamic loading tests reported by Ng (1989) for driven piles in loose granitic saprolites (with SPT N values less than 30) indicated that the measured capacities increased by 15% to 25% in the 24 hours after installation. The apparent 'set up' may have resulted from dissipation of positive excess pore water pressure generated during pile driving. As a general guideline, a driven pile should be tested at least three days after driving if it is driven into a granular material and at least four weeks after driving into a clayey soil, unless sufficient local experience or results of instrumentation indicate that a shorter period would be adequate for dissipation of excess pore pressure. 9.2.4.3 Static Pile Loading Tests

a) Reaction Arrangement To ensure stability of the test assembly setup, careful consideration should be given to the provision of a suitable reaction system. The geometry of the arrangement should also aim to minimise interaction between the test pile, reaction system and reference beam supports. It is advisable to have, say, a minimum of 20% margin on the capacity of the reaction against maximum test load.


March 2009 9-15

i) Compression tests Kentledge is commonly used in Malaysia as the reaction system (Figure 9.3). This involves the use of dead weights (compr ises of concrete blocks) supported by a deck of steel beams sitting on crib pads. The area of the crib should be sufficient to avoid bearing failure or excessive settlement of the ground. It is recommended that the crib pads are placed at least 1.3 m from the edge of the test pile to minimise interaction effects. If the separation distance is less than 1.3 m, the surcharge effect from the kentledge should be determined and allowed for in the interpretation of the loading test results. Sometimes tension piles are used to provide reaction for the applied load (Figure 9.4) and should be located as far as practicable from the test pile to minimise interaction effects. A minimum centre-to-centre spacing of 2 m or three pile diameters, whichever is greater, between the test pile and tension piles is recommended. If the centre spacing between piles is less than three pile diameters, there may be significant pile interaction and the observed settlement of the test pile will be less than what should have been. If a spacing of less than three pile diameters is adopted, uplift of the tension piles should be monitored and corrections should be made for the settlement of the test pile based on recognised methods considering pile interaction. A minimum of three reactions piles should be used to prevent instability of the set up during pile loading tests. Alternatively some from of lateral support should be provided.

Figure 9.3 Typical Arrangement of a Compression Test using Kentledge


9-16 March 2009

Figure 9.4 Typical Arrangement of a Compression Test using Tension Piles To reduce interaction between the ground anchors and the test pile, the fixed lengths of the anchors should be positioned a distance away from the centre of the test pile of at least three pile of diameters or 2 m, whichever is greater. Ground anchors may be used instead of tension piles to provide load reaction. The main shortcomings with ground anchors are the tendon flexibility and their vulnerability to lateral instability. The provision of a minimum of four ground anchors is preferred for safety considerations. Installation and testing of each ground anchor should be in accordance with the recommendations as given in GCO (1989) for temporary anchors. The anchor load should be locked off at 110% design working load. The movements of the anchor should be monitored during the loading tests to give prior warning of any imminent abrupt failure. The use of ground anchors will generally be most suitable in testing a raking pile because the horizontal component of the jacking may not be satisfactorily restrained in other reaction systems. They should be inclined along the same direction as the raking pile.


March 2009 9-17

Traditionally, a static loading test is carried out by jacking a pile against a kentledge or a reaction frame supported by tension piles or ground anchors. In recent years, Osterberg load cell (O-cell) has been widely adopted for static loading tests for large-diameter cast-in- place concrete piles. It can also be used in driven steel piles. An O-cell is commonly installed at or near the bottom of the pile. Reaction to the upward force exerted by the O-cell is provided by the shaft resistance. For such testing arrangement, the shaft resistance mobilised in the pile will be in upward direction. A smaller kentledge may be assembled in case the shaft resistance alone is not adequate to resist the applied load. The maximum test load is governed by either the available shaft resistance, the bearing stress at the base or the capacity of the O-cell itself.

ii) Uplift loading tests A typical arrangement for uplift loading tests is shown in Figure 9.4. The arrangement involving jacking at the centre is preferred because an even load can be applied

Reaction piles should be placed at least three test pile diameters, or a minimum of 2 m, from the centre of the test pile. Where the spacing is less than this, corrections for possible pile interaction should be made. Alternatively, an O-cell installed at the base of pile can also be used in an uplift test. iii) Lateral loading tests In a lateral loading test, two piles or pile groups may be jacked against each other (Figure 9.5(a)). It is recommended that the centre spacing of the piles should preferably be a minimum of ten pile diameters (CGS, 1992). Alternative reaction systems including a 'deadman' or weighted platform are also shown in Figure 9.5(b) and (c). 9.2.5 Equipment 9.2.5.1 Measurement of Load A typical load application and measurement system consists of hydraulic jacks, load measuring device, spherical seating and load bearing plates (Figure 9.3). The jacks used for the test should preferably be large-diameter low-pressure jacks with a travel of at least 15% of the pile diameter (or more if mini-piles are tested). A single jack is preferred where practicable. If more than one jack is used, then the pressure should be applied using a motorised pumping unit instead of a hand pump. Pressure gauges should be fitted to permit a check on the load. The complete jacking system including the hydraulic cylinder, valves, pump and pressure gauges should be calibrated as a single unit. It is strongly recommended that an independent load-measuring device in the form of a load cell, load column or pressure cell is used in a loading test. The device should be calibrated before each series of tests to an accuracy of not less than 2% of the maximum applied load. It is good practice to use a spherical seating in between the load measuring device and bearing plates in a compression loading test in order to minimise angular misalignment in the system and ensure that the load is applied coaxially to the test pile. Spherical seating is however only suitable for correcting relatively small angular misalignment of not more than about 3°.


9-18 March 2009

A load bearing plate should be firmly bedded onto the top of the pile (or the pile cap) orthogonal to the direction of applied load so as to spread the load evenly onto the pile. An O-cell consists of two steel plates between which there is an expandable pressurised chamber. Hydraulic fluid is injected to expand the chamber, which pushes the pile segment upward. At the same time, the bearing base (or lower pile segment if the O-cell is installed in middle of the pile) is loaded in the downward direction. Pressure gauges are attached to fluid feed lines to check the applied load and it is necessary to calibrate the O-cell. Correction may be needed to allow for the level difference between the pressure gauges, which is located at the ground surface and the load cell, which is usually installed at the base of the piles.

March 200

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9-20 March 2009

In a compression or tension test, measurements should be taken by four dial gauges evenly spaced along the perimeter of the pile to determine whether the pile head tilts significantly. The measuring points of the gauges should sit on the pile head or on brackets mounted on the side of the pile with a glass slide or machined steel plate acting as a datum for the stems. Care should be taken to ensure that the plates are perpendicular to the pile axis and that the dial gauge stems are in line with the axis. In a lateral loading test, dial gauges should be placed on the back of the pile with the stems in line with the load for measuring pile deflection (Figure 9.5). A separate system involving the use of a wire, mirror and scale may be used as a check on the dial gauges. The wire should be held under constant tension and supported from points at a distance not less than five pile diameters from the test pile and any part of the reaction system. Rotational and transverse movement of the pile should also be measured. LVDT can be used in place of dial gauges and readings can be taken remotely. However, they are susceptible to dirt and should be properly protected in a test. The reference beams to which the dial gauges or LVDT are attached should be rigid and stable. A light lattice girder with high stiffness in the vertical direction is recommended. This is better than heavy steel sections of lower rigidity. To minimise disturbance to the reference beams, the supports should be firmly embedded in the ground away from the influence of the loading system (say 2 m from piles or 1 m from kentledge support). It is recommended that the beam is clamped on one side of the support and free to slide on the other. Such an arrangement allows longitudinal movement of the beam caused by changes in temperature. The test assembly should be shaded from direct sunlight. In an axial loading test, levels of the test pile and reference beam supports should be monitored by an optical levelling system throughout the test to check for gross errors in the measurements. The optical levelling should be carried out at the maximum test load of each loading cycle and when the pile is unloaded at the end of each cycle. The use of precision levelling equipment with an accuracy of at least 1 mm is preferred. The datum for the optical levelling system should be stable and positioned sufficiently far away from the influence zone of the test. In loading tests using O-cell, rod extensometers are connected to the top and bottom plates of the O-cell (Figure 9.6). They are extended to the ground surface such that the movement of the plates can be measured by dial gauges or displacement transducers independently.

March 200

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9-22 March 2009

b) Maintained-load tests In a maintained-load test, the load is applied in increments, each being held until the rate of movement has reduced to an acceptably low value before the next load increment is applied. It is usual practice to include a number of loading and unloading cycles in a loading test. Such cycles can be particularly useful in assessing the onset of plastic movements by observing development of the residual (or plastic) movement with increase in load. Details of the common loading procedures used in Hong Kong GEO which can be used as a guide are summarised in Table 9.4. When testing a preliminary pile, the pile should, where practicable, be loaded to failure or at least to sufficient movement (say, a minimum of 5% of pile diameter). If the pile is loaded beyond 2 WL, a greater number of small load increments, of say 0.15 to 0.2 WL as appropriate, may be used in order that the load-settlement behaviour can be better defined before pile failure. However, the test load should not exceed the structural capacity of the pile. In principle, the same loading procedures suggested for compression tests may be used for lateral and uplift loading tests.


March 2009 9-23

Table 9.4 Loading Procedures and Acceptance Criteria for Pile Loading Tests in Hong Kong

General Specification for Civil Engineering Works (HKG, 1992)

Cycle 1-25% Qmax Cycle 2-50% Qmax Cycle 3-100% Qmax

1. δQ<2 x δ90%Q and 2. δ<2 mm for buildings

at working load and 10 mm for other structures (e.g. bridges) at working load.

1. Load increments/ decrements to be in 25% of the design working load; pile to be unloaded at the end of each cycle.

2. Preliminary piles are to be tested to not less than twice the design working load (i.e. Qmax>2WL); working piles to be tested to not less than 1.8 times design working load (i.e. Qmax>1.8WL).

3. Load increments/ decrements not to be applied until rate of settlement or rebound of pile is less than 0.1 mm in 20 minutes.

4. Full load at each cycle to be maintained for at least 24 hours after rate of settlement has reduced to less than 0.1 mm per hour.

Code of Practice for Foundations (BD, 2004a)

Loading schedule for piles with a diameter or at least lateral dimension not exceeding 750 mm: Cycle 1 – 100% WL Cycle 2 – 200% WL (=Qmax)

1. δmax<Q LA E

D 4

(mm) 2. The greater of: δmax<

D 4 or 0.25δmax (in mm)

1. Load increments/ decrements to be in 50% of the design working load; pile to be unloaded at the end of each cycle.

2. Piles are to e tested to twice design working load.

3. Increments of load not to be applied until rate of settlement or recovery of pile is less than 0.05 mm in 10 minutes.

4. Full load at cycle 2 should be maintained for at least 72 hours.

5. The residual settlement, δres, should be taken when the rate of recovery of the pile after removal of test load is less than 0.1 mm in 15 minutes.

Legend : δQ = pile head settlement at failure or maximum test load δ90%Q = pile head settlement at 90% of failure or maximum test load

δmax = maximum pile head settlement δ = pile head settlement


9-24 March 2009

δres = residual (or permanent) pile head settlement upon unloading from maximum

load Qmax = maximum test load WL = design working load of pile L = pile length Ap = cross-sectional area of pile Ep = Young's modulus of pile D = least lateral dimension of pile section (mm)

9.2.5.4 Instrumentation a) General Information on the load transfer mechanism can be derived from a loading test if the pile is instrumented. To ensure that appropriate and reliable results can be obtained, the pile instrumentation system should be compatible with the objectives of the test. Important aspects including selection, disposition and methods of installation should be carefully considered. It is essential that sufficient redundancy is built in to allow for possible damage and malfunctioning of instruments. Where possible, isolated measurements i.e., survey leveling method should be made using more than one type of equipment to permit cross-checking of results. An understanding of the ground profile, proposed construction technique and a preliminary assessment of the probable behaviour of the pile will be helpful in designing the disposition of the instruments. Limitations and resolutions of the instruments should be understood. b) Axial loading tests Information that can be established from an instrumented axial loading test includes the distribution of load and movement, development of shaft resistance and end-bearing resistance with displacement. A typical instrumentation layout is given in Figure 9.6. Strain gauges (electrical resistance and vibrating wire types) can be used to measure local strains, which can be converted to stresses or loads. Vibrating wire strain gauges are generally preferred, particularly for long-term monitoring, as the readings will not be affected by changes in voltage over the length of cable used, earth leakage, corrosion to connection and temperature variation. In case measurements need to be taken rapidly, e.g. in simulation dynamic response of piles, electrical resistance type strain gauges are more suitable. A variant form of vibrating wire strain gauges is the 'sister bar' or 'rebar strain meter'. This is commonly used in cast-in-place concrete piles. It consists of a vibrating strain gauge assembled inside a high strength steel housing that joins two reinforcement bars at both ends by welding or couplers. The sister bar can replace a section of the steel in the reinforcement cage or be placed alongside it. Such an arrangement minimises the chance that a strain gauge is damaged during placing of concrete. The electrical wirings should be properly tied to the reinforcement cage at regular intervals. To measure axial loads, the strain gauge stems are orientated in line with the direction of the load (i.e. vertical gauges). One set of gauges should be placed near the top of the pile, and preferably in a position where the pile shaft is not subject to external shaft resistance, to facilitate calculation of the modulus of the composite section. Gauges should also be placed close to the base of the


March 2009 9-25

pile (practically 0.5 m) with others positioned near stratum boundaries and at intermediate levels. A minimum of two and preferably four gauges should be provided at each level where practicable. c) Lateral loading tests The common types of internal instrumentation used in a lateral loading test are inclinometers, strain gauges and electro-levels. The deflected shape of a pile subject to lateral loading can be monitored using an inclinometer. The system consists of an access tube and a torpedo sensor. For cast-in-place piles, the tube is installed in the pile prior to concreting. For displacement piles such as H- piles, a slot can be reserved in the pile by welding on a steel channel or angle section prior to pile driving. The tube is grouted into the slot after driving. During the test, a torpedo is used to measure the slope, typically in 0.5 m gauge lengths, which can be converted to deflections. Care needs to be exercised in minimising any asymmetrical arrangement of the pile section or excessive bending of the pile during welding of the inclinometer protective tubing. In extreme cases, the pile may become more prone to being driven off vertical because of these factors. Strain gauges with their stems orientated in line with the pile axis can be used for measuring direct stresses and hence bending stresses in the pile. They can also be oriented horizontally to measure lateral stresses supplemented by earth pressure cells. Electro-levels measure changes in slope based on the inclination of an electrolytic fluid that can move freely relative to three electrodes inside a sealed glass tube (Price & Wardle, 1983; Chan & Weeks, 1995). The changes in slope can be converted to deflections by multiplying the tangent of the change in inclination by the gauge length. The devices are mounted in an inclinometer tube cast into the pile and can be replaced if they malfunction after installation. Earth pressure cells can also be used to measure the changes in normal stresses acting on the pile during loading. It is important that these pressure cells are properly calibrated for cell action factors, etc. to ensure sensible results are being obtained. 9.2.5.5 Interpretation of Test Results a) General A considerable amount of information can be derived from a pile loading test, particularly with an instrumented pile. In the interpretation of test results for design, it will be necessary to consider any alterations to the site conditions, such as fill placement, excavation or dewatering, which can significantly affect the insitu stress level, and hence the pile capacity, after the loading test. b) Evaluation of failure load Typical load-settlement curves, together with some possible modes of failure, are shown in Figure 9.7. Problems such as presence of a soft clay layer, defects in the pile shaft and poor construction techniques may be deduced from the curves where a pile has been tested to failure.

9-26

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March 2009 9-27

The acceptance criteria specified in the Code of Practice for Foundations (BD, 2004a) are generally adopted for engineering practice in Malaysia. Non-compliance with the criterion on acceptance criteria does not necessarily imply non-acceptance of the pile. Where this criterion is not met, it is prudent to examine the pile behaviour more closely to find out the reasons of non-compliance. In principle, a designer should concentrate on the limiting deflection at working load as well as the factor of safety against failure or sudden gross movements. The limiting settlement of a test pile at working load should be determined on an individual basis taking into account the sensitivity of the structure, the elastic compression component, effects of pile group interaction under working condition, and expected behaviour of piles as observed in similar precedents. In analysing the settlement behaviour of the pile under a pile loading test, it is worth noting that the applied load will be carried in part or entirely by the shaft resistance, although the shaft resistance may be ignored in the pile design. Consequently, the elastic compression component of pile could be smaller than that estimated based on the entire length of the pile, particularly for long friction pile. Fraser & Ng (1990) suggested that upon removal of the maximum test load, the recovery of the pile head settlement may be restricted by the 'locked in' stress as a result of reversal of shaft resistance upon removal of the test load. 9.2.6 Dynamic Loading Tests 9.2.6.1 General Various techniques for dynamic loading tests are now available. These tests are relatively cheap and quick to carry out compared with static loading tests. Information that can be obtained from a dynamic loading test includes: (a) static load capacity of the pile, (b) energy delivered by the pile driving hammer to the pile, (c) maximum driving compressive stresses (tensile stress should be omitted), (d) location and extent of structural damage. 9.2.6.2 Test Methods The dynamic loading test is generally carried out by driving a prefabricated pile or by applying impact loading on a cast-in-place pile by a drop hammer. A standard procedure for carrying out a dynamic loading test is given in ASTM (1995b). The equipment required for carrying out a dynamic pile loading test includes a driving hammer, strain transducers and accelerometers, together with appropriate data recording, processing and measuring equipment. The hammer should have a capacity large enough to cause sufficient pile movement such that the resistance of the pile can be fully mobilised. A guide tube assembly to ensure that the force is applied axially on the pile should be used. The strain transducers contain resistance foil gauges in a full bridge arrangement. The accelerometers consist of a quartz crystal which produces a voltage linearly proportional to the acceleration. A pair of strain transducers and accelerometers are fixed to opposite sides of the pile, either by drilling and bolting directly to the pile or by welding mounting blocks, and


9-28 March 2009

positioned at least two diameters or twice the length of the longest side of the pile section below the pile head to ensure a reasonably uniform stress field at the measuring elevation. In the test, the strain and acceleration measured at the pile head for each blow are recorded. The signals from the instruments are transmitted to a data recording, filtering and displaying device to determine the variation of force and velocity with time.

9.2.6.3 Methods of Interpretation a) General Two general types of analysis based on wave propagation theory, namely direct and indirect methods, are available. Direct methods of analysis apply to measurements obtained directly from a (single) blow, whilst indirect methods of analysis are based on signal matching carried out on results obtained from one or several blows. Examples of direct methods of analysis include CASE, IMPEDANCE and TNO method, and indirect methods include CAPWAP, TNOWAVE and SIMBAT, CASE and CAPWAP analyses are used mainly for displacement piles, although in principle they can also be applied to cast-in-place piles. SIMBAT has been developed primarily for cast-in- place piles, but it is equally applicable to displacement piles. In a typical analysis of dynamic loading test, the penetration resistance is assumed to be comprised of two parts, namely a static component, Rs, and a dynamic component, Rd. b) CAPWAP method CAPWAP (CAse Pile Wave Analysis Program) analysis is the common analysis adopted by the local tester in Malaysia. In a CAPWAP analysis, the soil is represented by a series of elasto-plastic springs in parallel with a linear dashpot similar to that used in the wave equation analysis proposed by Smith (1912). The soil can also be modelled as a continuum when the pile is relatively short. CAPWAP measures the acceleration-time data as the input boundary condition. The program computes a force versus time curve which is compared with the recorded data. If there is a mismatch, the soil model is adjusted. This iterative procedure is repeated until a satisfactory match is achieved between the computed and measured force-time diagrams. The dynamic component of penetration resistance is given by: Rd = jsvpRs (9.14) Where:

js = Smith damping coefficient vp = velocity of pile at each segment Rs = static component of penetration resistance

Input parameters for the analysis include pile dimensions and properties, soil model parameters including the static pile capacity, Smith damping coefficient, js and soil quake (i.e. the amount of elastic deformation before yielding starts), and the signals measured in the field. The output will be in the form of distribution of static unit shaft resistance against depth and base response, together with the static load-settlement relationship up to about 1.5 times the working load. It should be noted that the analysis does not model the onset of pile failure correctly and care should be exercised when predicting deflections at loads close to the ultimate pile capacity.


March 2009 9-29

9.2.6.4 Recommendations on the Use of Dynamic Loading Tests Traditionally, pile driving formulae are used as a mean to assess pile capacity from a measurement of 'set per blow' and are supplemented with static loading tests on selected piles. Although such an approach is the standard in local practice for driving piles, driving formulae are considered fundamentally incorrect and quantitative agreement between static pile capacities predicted by driving formulae and actual values cannot be relied upon. Dynamic load testing is preferred for pile capacity predictions. Dynamic load testing can be applied to non-homogeneous soils or piles with a varying cross-sectional area. The static load-settlement response of a pile can also be predicted. Dynamic pile loading tests can supplement the design of driven piles provided that they have been properly calibrated against static loading tests and an adequate site investigation has been carried out. It should be noted that such calibration of the analysis model has to be based on static loading tests on piles of similar length, cross section and under comparable soil conditions and loaded to failure. A static loading test, which is carried out to a proof load, is an inconclusive result for assessing the ultimate resistance of the pile. The reliability of the prediction of dynamic loading test methods is dependent on the adequacy of the wave equation model and the premise that a unique solution exists when the best fit is obtained within the limitation of the assumption of an elasto/rigid plastic soil behavior. In addition, there are uncertainties with the modelling of effects of residual driving stresses in the wave equation formulation. 9.3 LATERALLY LOADED PILES 9.3.1 Introduction The lateral load capacity of a pile may be limited by the following: (a) Shear capacity of the soil; (b) Structural (i.e. bending moment and shear) capacity of the pile section itself; and (c) Excessive deformation of the pile.

The failure mechanisms of short piles under lateral loads as compared to those of long piles differ, requiring therefore different and appropriate design methods. In order to establish if a pile behaves a rigid unit (i.e. short pile) or as a flexible member (i.e. long pile), the stiffness factors as defined in Figure 9.8 below will employed.


9-30 March 2009

Notes: 1. For constant soil modulus with depth (e.g. stiff overconsolidated clay), pile stiffness factor

R = EpIpkhD

4 (in units of length) where EpIp is the bending stiffness of the pile, D is the

width of the pile, kh is the coefficient of horizontal subgrade reaction.

2. For soil modulus increases linearly with depth (e.g. normally consolidated clay & granular

soils), pile stiffness factor, T = EpIpnh

5 where nh is the constant of horizontal subgrade

reaction given in table below:

3. The criteria for behaviour as a short (rigid) pile or as a long (flexible) pile are as follows:

Pile Type Soil Modulus

Linearly increasing Constant Short (rigid) piles Long (flexible) piles

L 2T L 4T

L 2R L 3.5R

Figure 9.7 Failure Modes of Vertical Piles under Lateral Loads (Broms, 1914a)


March 2009 9-31

Table 9.5 Typical Values of Coefficient of Horizontal Subgrade Reaction

Consistency

(MN/m3) Loose

(N value 4-Medium Dense

(N value 11-Dense

(N value 31-nh for dry or 2.2 6.6 17.6 nh for 1.3 4.4 10.7

Notes:

i. The above nh values are based on Terzaghi (1955) and are valid for stresses up to about half the ultimate bearing capacity with allowance made for long-term movements.

ii. For sands, Elson (1984) suggested that Terzaghi's values should be used as a lower limit and the

following relationship as the upper limits :

nh =

where Dr is the relative density of sand in percent.

iii. Other observed values of nh, which include an allowance for long-term movement, are as follows (Tomlinson, 1994) :

Soft normally consolidated clays: 350 to 700 Soft organic silts: 150 kN/m3

iv. For sands, nh may be related to the drained horizontal Young modulus (Eh') in MPa as follows

(Yoshida & Yoshinaka, 1972; Parry, 1972) :

nh = 0.8h

'to 1.8h'

z (9.16)

where z is depth below ground surface in metres. v. It should be noted that empirical relationships developed for transported soils between N value

and relative density are not generally valid for weathered rocks. Corestones, for example, can give misleading high values that are unrepresentative of the soil mass.

As the surface soil layer can be subject to disturbance, suitable allowance should be made in the design by ignoring as appropriate, the resistance of the upper part of the soil. 9.3.2 Lateral Load Capacity of Pile In respect of the ultimate lateral resistance of a c'- φ' material, the method proposed for short rigid piles by Brinch Hansen (1911) can be referred (Figure 9.9).


9-32 March 2009

Notes: 1. The above passive pressure coefficients Kqr and Kcz are obtained based on the method

proposed by Brinch Hansen (1961). Unit passive resistance per unit width, pz at depth z is: pz = σv’ Kqz + c’Kcz (9.17)

where σv’ is the effective overburden pressure at depth z, c’ is the apparent cohesion of soil at depth z.


March 2009 9-33

2. The point of rotation (Point X) is the point at which the sum of the moment (∑ M) of the

passive pressure about the point of application of the horizontal load is zero. This point can be determined by a trial and adjustment process.

∑ M= ∑ pz

L

ne1+z D- ∑ pz

z=Lz=x

z=hz=0

L

ne1+z D (9.18)

3. The ultimate lateral resistance of a pile to the horizontal force Hu can be obtained by taking

moment about the point rotation, i.e. Hu e1+x = ∑ pz

L

nD x-z - ∑ pz

z=Lz=x

z=hz=0

L

nz-x D (9.19)

4. An applied moment M can be replaced by a horizontal force H at a distance e1 above the

ground surface where M = H e1.

5. When the head of a pile is fixed against rotation, the equivalent height, eo above the point of fixity of a force H acting on a pile with a free-head is given by eo = 0.5 (e1 + zf) is the depth from the ground surface to point of virtual fixity. ACI (1980) recommended that zf should be taken as 1.4R for stiff, overconsolidated clays and 1.8T for normally consolidatedclays, granular soils and silts, and peat. Pile stiffness factors, R and T, can ve determined based on Figure.

Figure 9.8 Coefficients Kqz and Kcz at Depth z for Short Piles Subject to Lateral Load (Brinch Hansen,

1911)

Methods of calculating the ultimate lateral soil resistance for fixed-head and free-head piles in granular soils and clays are put forward by Broms (1914a & b). The theory is similar to that of Brinch Hansen except that some simplifications are made in respect of the distribution of ultimate soil resistance with depth. The design for short and long piles in granular soils are summarised in Figures 9.10 and 9.11 respectively. Kulhawy & Chen (1992) compared the results of a number of field and laboratory tests on bored piles. They found that Brom’s method tended to underestimate the ultimate lateral load by about 15% to 20%.


9-34 March 2009

Figure 9.9 Ultimate Lateral Resistance of Short Piles in Granular Soils (Broms, 1914a)

Notes:

1. For free-head short piles in granular soils 2.

Hu= 0.5 DL3Kp s'

e1+L

Where Kp = Rankine’s coefficient of passive pressure = 1+sin '

1- sin '

D = width of the pile Ø’ = angle of shearing resistance of soil s = effective unit weight of soil

3. For fixed-head short piles in granular soils

4. Hu = 1.5 DL2KP ρ C’ The above equation is valid only when the maximum bending moment, Mmax develops at the pile head is less than the ultimate moment of resistance, Mu, of the pile at this point. The bending moment is given by Mmax = DL3KP ρ C’

5. PL is the concentrated horizontal force at pile tip due to passive soil resistance.


March 2009 9-35

Figure 9.10 Ultimate Lateral Resistance of Long Piles in Granular Soils (Broms, 1914b)

D = width of the pile in the direction of Ø’ = angle of shearing resistance s’ = effective unit weight of soil Kp = Rankine’s coefficient of passive

Notes: 1. For free-head long piles in granular soils, Mmax = H(e1+0.67f*)

where f* = 0.82 H

s'DKp

2. For fixed-headed short piles in granular soils, the maximum bending moment

occurs at the pile head and at the ultimate load. It is equal to the ultimate moment of resistance of pile shaft.

Mmax = 0.5H (e1+0.67f*) For a pile of uniform cross-section, the ultimate value of lateral load Hu is given by taking Mmax as the ultimate moment of resistance of the pile, Mu.


9-36 March 2009

Poulos (1985) has extended Broms' methods to consider the lateral load capacity of a pile in a two-layer soil. The design approaches presented above are simplified representations of the pile behaviour. Nevertheless, they form a useful framework for obtaining a rough estimate of the likely capacity, and experience suggests that they are generally adequate for routine design. In situations where the design is likely to be governed by lateral load behaviour, loading tests should be carried out to justify the design approach and verify the design parameters. The bending moment and shearing force in a pile subject to lateral loading may be assessed using the method by Matlock & Reese (1910) as given in Figures 9.12 and 9.13. The tabulated values of Matlock & Reese have been summarised by Elson (1984) for easy reference. This method models the pile as an elastic beam embedded in a homogeneous or non-homogeneous soil. In long, flexible piles, the structural capacity is likely to govern the ultimate capacity of a laterally-loaded pile. Relatively short less than critical length given in Figure 9.8 end-bearing piles, e.g. piles founded on rock, with toe being effectively fixed against both translation and rotation, can be modelled as cantilevers cast at the bottom, with the top either fixed or free, depending on restraints on pile head. Accordingly, the lateral stiffness of the overburden can thus be represented by springs with appropriate stiffness.


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Figure 9.11 Influence Coefficients for Piles with Applied Lateral Load and Moment (Flexible Cap or

Hinged End Conditions) (Matlock & Reese, 1910)

Deflection Coefficient, Fs for Applied Moment M Deflection Coefficient, Fs for Applied Lateral Load, H

Moment Coefficient, FM for Applied Moment M Moment Coefficient, FM for Applied Lateral Load, H

Shear Coefficient, Fv for Applied Moment M Shear Coefficient, Fv for Applied Lateral Load, H

Notes:

1. T = EpIpnh

5 where EpIp = bending stiffness of pile and nh = constant of

horizontal subgrade reaction 2. 3. Obtain coefficients Fδ,FM and Fv at appropriate depths desired and

compute deflection, moment and shear respectively using the given formulae.


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Figure 9.12 Influence Coefficients for Piles with Applied Lateral Load (Fixed against Rotation at Ground Surface) (Matlock & Reese, 1910)

Deflection Coefficient, Fδ for Applied Lateral Load, H

Moment Coefficient, FM for Applied Lateral Force, H

Notes:

1. T = EpIpnh

5 where EpIp = bending stiffness of pile and nh = constant of horizontal

subgrade reaction

2. Obtain coefficients Fδ,FM and Fv at appropriate depths desired and compute deflection, moment and shear respectively using the given formulae.

3. Maximum shear occurs at top of pile and is equal to the applied load H.


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The minimum factors of safety recommended for design are summarised in Table 9.3. For vertical piles designed to resist lateral load, it is usually governed by the limiting lateral deflection requirements. For piles in sloping ground, the ultimate lateral resistance can be affected significantly if the piles are positioned within a distance of about five to seven pile diameters from the slope crest. Based on full-scale test results, Bhushan et al (1979) proposed that the lateral resistance for level ground be factored by 1/(1 + tan θs), where θs is the slope angle. Alternatively, Siu (1992) proposed a simplifying method for determining the lateral resistance of a pile in sloping ground taking into account three-dimensional effects. 9.3.3 Inclined Loads If a vertical pile is subjected to an inclined and eccentric load, the ultimate bearing capacity in the direction of the applied load is intermediate between that of a lateral load and a vertical load because the passive earth pressure is increased and the vertical bearing capacity is decreased by the inclination and eccentricity of the load. Based on model tests, Meyerhof (1981) suggested that the vertical component Qv, of the ultimate eccentric and inclined load can be expressed in terms of a reduction factor rf on the ultimate concentric vertical load Qo, as given in Figure 9.13. The lateral load capacity can be estimated following the methods given in Item 9.3.2 above. Piles, subjected to inclined loads, should also be checked against possible buckling, pile head deflection and induced bending moments. 9.3.4 Raking Piles in Soil Raking piles provide a common method of resisting lateral loads. For the normal range of inclination of raking piles used in practice, the raking pile may be considered as an equivalent vertical pile subjected to inclined loading. Deformations and forces induced in a general pile group comprising vertical and raking piles under combined loading condition are not amenable to presentation in graphical or equation format. A detailed analysis will invariably require the use of a computer. Zhang et al (2002) conducted centrifuge tests to investigate the effect of vertical load on the lateral response of a pile group with raking piles. The results of the experiments indicated that there was a slight increase in the lateral resistance of the pile groups with the application of a vertical load. a) Methodologies for Analysis i) Stiffness method can be used to analyse pile groups comprising vertical piles and raking piles installed to any inclination. In this method, the piles and pile cap form a structural frame to carry axial, lateral and moment loading. The piles are assumed to be pin-jointed and deformed elastically. The load on each pile is determined based on the analysis of the structural frame. The lateral restraint of the soil is neglected and this model is not a good representation of the actual behaviour of the pile group. The design is inherently conservative and other forms of analyses are preferred for pile groups subjected to large lateral load and moment (Elson, 1984). ii) A more rational approach is to model the soil as an elastic continuum. A number of commercial computer programs have been written for general pile group analysis based on idealising the soil as a linear elastic material, e.g. PIGLET (Randolph, 1980), DEFPIG (Poulos, 1990a), PGROUP (Bannerjee & Driscoll, 1978). The first two programs are based on the interaction factor method


9-40 March 2009

while the last one uses the boundary element method. A brief summary of the features of some of the computer programs developed for analysis of general pile groups can be found in Poulos (1989b) and the report by the Institution of Structural Engineers (ISE, 1989). Computer analyses based on the elastic continuum method generally allow more realistic boundary conditions, variation in pile stiffness and complex combined loading to be modelled.

Comparisons between results of different computer programs for simple problems have been carried out, e.g. O'Neill & Ha (1982) and Poulos & Randolph (1983). The comparisons are generally favourable with discrepancies which are likely to be less than the margin of uncertainty associated with the input parameters. Comparisons of this kind lend confidence in the use of these programs for more complex problems.

Pile group analysis programs can be useful to give an insight into the effects of interaction and to provide a sound basis for rational design decisions. In practice, however, the simplification of the elastic analyses, together with the assumptions made for the idealisation of the soil profile, soil properties and construction sequence could potentially lead to misleading results for a complex problem. Therefore, considerable care must be exercised in the interpretation of the results.

The limitations of the computer programs must be understood and the idealisations and assumptions made in the analyses must be compatible with the problem being considered. It would be prudent to carry out parametric studies to investigate the sensitivity of the governing parameters for complex problems. b) Choice of Parameters One of the biggest problems faced by a designer is the choice of appropriate soil parameters for analysis. Given the differing assumptions and problem formulation between computer programs, somewhat different soil parameters may be required for different programs for a certain problem. The appropriate soil parameters should ideally be calibrated against a similar case history or derived from the back analysis of a site-specific instrumented pile test using the proposed computer program for a detailed analysis. 9.3.5 Lateral Loading 9.3.5.1 General The response of piles to lateral loading is sensitive to soil properties near the ground surface. Due to the proneness to disturbance of these surface layers, reasonably conservative soil parameters should be adopted in the prediction of pile deflection. An approximate assessment of the effects of soil layering can be made by reference to the work by Davisson & Gill (1913) or Pise (1982). Poulos (1972) studied the behaviour of a laterally-loaded pile socketed in rock. He concluded that socketing of a pile has little influence on the horizontal deflection at working load unless the pile is sufficiently rigid, with a stiffness factor under lateral loading, Kr, greater than 0.01, where

Kf = EpIpEsL

4 (9.20)

and Ip and L are the second moment of area and length of the pile respectively. The effect of sloping ground in front of a laterally-loaded pile was analysed by Poulos (1971) for clayey soils, and by Nakashima et al (1985) for granular soils. It was concluded that the effect on pile deformation will not be significant if the pile is beyond a distance of about five (5) to seven (7) pile diameters from the slope crest.


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The load-deflection and load-rotation relationships for a laterally-loaded pile are generally highly non-linear. Three approaches have been proposed for predicting the behaviour of a single pile: (a) The equivalent cantilever method, (b) The subgrade reaction method, and (c) The elastic continuum method. Alternative methods include numerical methods such as the finite element and boundary element methods as discussed in the subsequent sections of this chapter. However, these are seldom justified for routine design problems. A useful summary of the methods of determining the horizontal soil stiffness is given by Jamiolkowski & Garassino (1977). It should be noted that the currently available analytical methods for assessing deformation of laterally-loaded piles do not consider the contribution of the side shear stiffness. Some allowance may be made for barrettes loaded in the direction of the long side of the section with the use of additional springs to model the shear stiffness and capacity in the subgrade reaction approach. Where the allowable deformation is relatively large, the effects of non-linear bending behaviour of the pile section due to progressive yielding and cracking, along with its effect on the deflection and bending moment profile should be considered (Kramer & Heavey, 1988). The possible non-linear structural behaviour of the section can be determined by measuring the response of an upstand above the ground surface in a lateral loading test. 9.3.5.2 Equivalent Cantilever Method This method represents a gross simplification of the problem and should only be used as an approximate check on the other more rigorous methods unless the pile is subject to nominal lateral load. In this method, the pile is represented by an equivalent cantilever and the deflection is computed for either free-head or fixed-head conditions. Empirical expressions for the depths to the point of virtual fixity in different ground conditions are summarised by Tomlinson (1994). The principal shortcoming of this approach is that the relative pile-soil stiffness is not considered in a rational framework in determining the point of fixity. Also, the method is not suited for evaluating profiles of bending moments. 9.3.5.3 Subgrade Reaction Method In this method, the soil is idealised as a series of discrete springs down the pile shaft. The continuum nature of the soil is not taken into account in this formulation. The characteristic of the soil spring is thus expressed as follows: p = kh δh (9.21) Ph = Kh δh (9.22) = kh D δh (for constant Kh) = nh z δh (for the case of Kh varying linearly with depth) Where: p = soil pressure kh = coefficient of horizontal subgrade reaction δh = lateral deflection


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Ph = soil reaction per unit length of pile Kh = modulus horizontal subgrade reaction D = width or diameter of pile nh = constant of horizontal subgrade reaction, sometimes referred to as the constant of

modulus variation in the literature z = depth below ground surface It should be noted that kh is not a fundamental soil parameter as it is influenced by the pile dimensions. In contrast, Kh is more of a fundamental property and is related to the Young's modulus of the soil, and it is not a function of pile dimensions. Soil springs determined using subgrade reaction do not consider the interaction between adjoining springs. Calibration against field test data may be necessary in order to adjust the soil modulus to derive a better estimation (Poulos et al, 2002). Traditionally, over-consolidated clay is assumed to have a constant Kh with depth whereas normally consolidated clay and granular soil is assumed to have a Kh increasing linearly with depth, starting from zero at ground surface. For a uniform pile with a given bending stiffness (EpIp), there is a critical length (Lc) beyond which the pile behaves as if it were infinitely long and can be termed a 'flexible' pile, under lateral load. The expressions for the critical lengths are thus given as follows:

Lc = 4EpIpKh

4 (9.23)

= 4 R for soils with a constant Kh

Lc = 4EpIpnh

5 (9.24)

= 4 T for soils with a Kh increasing linearly with depth The terms 'R' and 'T' are referred to as the characteristic lengths by Matlock & Reese (1910) for homogeneous soils and non-homogeneous soils, respectively. They derived generalised solutions for piles in granular soils and clayey soils. The solutions for granular soils as summarized in Figures 9.12 and 9.13. A slightly different approach has been proposed by Broms (1914a & b) in which the pile response is related to the parameter L/R for clays, and to the parameter L/T for granular soils. The solutions provide the deflection and rotation at the head of rigid and flexible piles. In general, the subgrade reaction method can give satisfactory predictions of the deflection of a single pile provided that the subgrade reaction parameters are derived from established correlations or calibrated against similar case histories or loading test results. Typical ranges of values of nh, together with recommendations for design approach, are given in Table 9.5, previously. The parameter kh can be related to results of pressuremeter tests (CGS, 1992). The effects of pile width and shape on the deformation parameters are discussed by Siu (1992). The solutions by Matlock & Reese (1910) apply for idealised, single layer soil. The subgrade reaction method can be extended to include non-linear effects by defining the complete load transfer curves or 'p-y' curves. This formulation is more complex and a nonlinear analysis generally requires the use


March 2009 9-43

of computer models similar to those described by Bowles (1992), which can be used to take into account variation of deformation characteristics with depth. In this approach, the pile is represented by a number of segments each supported by a spring, and the spring stiffness can be related to the deformation parameters by empirical correlations (e.g. SPT N values). Due allowance can and should be made for the strength of the upper, and often weaker, soils whose strength may be fully mobilised even at working load condition. Alternatively, the load-transfer curves can be determined based on instrumented pile loading tests, in which a series of 'p-y' curves are derived for various types of soils. Nip & Ng (2005) presented a simple method to back-analyse results of laterally loaded piles for deriving the 'p-y' curves for superficial deposits. Reese & Van Impe (2001) discussed factors that should be considered when formulating the 'p-y' curves. These include pile types and flexural stiffness, duration of loading, pile geometry and layout, effect of pile installation and ground conditions. Despite the complexities in developing the 'p-y' curves, the analytical method is simple once the non-linear behaviours of the soils are modelled by the 'p-y' curves. This method is particularly suitable for layered soils. 9.3.5.4 Elastic Continuum Method Solutions for deflection and rotation based on elastic continuum assumptions are summarised by Poulos & Davis (1980). Design charts are given for different slenderness ratios (L/D) and the dimensionless pile stiffness factors under lateral loading (Kr) for both friction and end-bearing piles. The concept of critical length is however not considered in this formulation as pointed out by Elson (1984). A comparison of these simplified elastic continuum solutions with those of the rigorous boundary element analyses have been carried out by Elson (1984). The comparison suggests that the solutions by Poulos & Davis (1980) generally give higher deflections and rotations at ground surface, particularly for piles in a soil with increasing stiffness with depth. The elastic analysis has been extended by Poulos & Davis (1980) to account for plastic yielding of soil near ground surface. In this approximate method, the limiting ultimate stress criteria as proposed by Broms (1915) have been adopted to determine factors for correction of the basic solution. An alternative approach is proposed by Randolph (1981b) who fitted empirical algebraic expressions to the results of finite element analyses for homogeneous and non-homogeneous linear elastic soils. In this formulation, the critical pile length, Lc (beyond which the pile plays no part in the behaviour of the upper part) is defined as follows:

Lc = 2roEpe

Gc

27

⁄ (9.25)

Where: G* = G(1+0.75vs) Gc = mean value of G* over the critical length, Lc, in a flexible pile G = shear modulus of soil ro = radius of an equivalent circular pile vs = Poisson’s ration of soil EpIp = bending stiffness of actual pile


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Epe = equivalent Young’s modulus of the pile = 4EpIp

ro4

For a given problem, iterations will be necessary to evaluate the values of Lc and Gc. Expressions for deflection and rotation at ground level given by Randolph's elastic continuum formulation are summarised in Figure 9.14. Results of horizontal plate loading tests carried out from within a hand-dug caisson in completely weathered granite (Whiteside, 1981) indicate the following range of correlation: Eh' = 0.1 N to 1.9 N (MPa) (9.26) where Eh' is the drained horizontal Young's modulus of the soil. The modulus may be nearer the lower bound if disturbance due to pile excavation and stress relief is excessive. The reloading modulus was however found to be two to three times the above values. Plumbridge et al (2000b) carried out lateral loading tests on large-diameter bored piles and barrettes in fill and alluvial deposits. Testing arrangement on five sites included a 100 cycle bi-directional loading stage followed by a five-stage maintained lateral loading test. The cyclic loading indicated only a negligible degradation in pile-soil stiffness after the 100 cycle bi-direction loading. The deflection behaviour for piles in push or pull directions was generally similar. Based on the deflection profile of the single pile in maintained-load tests, the correlation between horizontal Young's modulus, Eh' and SPT N value was found to range between 3 N and 4 N (MPa). Lam et al (1991) reported results of horizontal Goodman Jack tests carried out from within a caisson in moderately to slightly (Grade III / II) weathered granite. The interpreted rock mass modulus was in the range of 3.1 to 8.2 GPa. In the absence of site-specific field data, the above range of values may be used in preliminary design of piles subject to lateral loads.

Figure 9.13 Analysis of Behaviour of a Laterally Loaded Pile Using the Elastic Continuum Method (Randolph, 1981a)

Free-head Piles

δh = Ep/Gc

17

ρc'Gc

0.27H0.5Lc

+ 0.3M0.5Lc 2

Ө = Ep/Gc

17

ρc'Gc

0.3H.

+0.8 'M

0.5Lc 3

The maximum moment for a pile under a lateral load H occurs at depth between 0.25Lc (for homogenous soil) and 0.33Lc (for soil with stiffness proportional to depth). The value of the maximum bending moment Mmax may be approximated using the following expression:

Mmax = 0.1ρc'

H Lc


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In this case, the pile rotation at ground surface, Ө, equals zero and the fixing moment, Mf, and lateral deflection, δh, are given by the following expression:

Mf = -0.375H(o.5Lc)

ρc' (9.27)

δh = (Ep/Gc)

17

ρc'Gc0.27-

0.11

ρc'

H0.5Lc

(9.28)

Where: δh = lateral pile deflection at ground surface Ө = pile rotation at ground surface Gc = characteristic shear modulus, i.e. average value of G* over the critical length Lc of

the pile

Lc = critical pile length for lateral loading = 2ro EG

Epo = equivalent Young’s modulus of pile = 4EpIp

ro4

c’ = degree of homogeneity over critical length, Lc = G*0.25Lc

Gc

G* = G(1+0.75vs) G*0.25Lc = value of G* at depth of 0.25Lc vc = Poisson’s ratio of soil G = shear modulus of soil H = horizontal load M = bending moment EpIp = bending stiffness of pile ro = pile radius The lateral deflection of a fixed-head pile is approximately half that of a corresponding free-head pile. 9.4 PILE GROUP 9.4.1 General Piles installed in a group to form a foundation will, when loaded, give rise to interaction between individual piles as well as between the structure and the piles. The pile- soil-pile interaction arises as a result of overlapping of stress (or strain) fields and could affect both the capacity and the settlement of the piles. The piled foundation as a whole also interacts with the structure by virtue of the difference in stiffness. This foundation-structure interaction affects the distribution of loads in the piles, together with forces and movements experienced by the structure. The analysis of the behaviour of a pile group is a complex soil-structure interaction problem. The behaviour of a pile group foundation will be influenced by, inter alia: (a) Method of pile installation, e.g. replacement or displacement piles, (b) Dominant mode of load transfer, i.e. shaft resistance or end- bearing, (c) Nature of founding materials, (d) Three-dimensional geometry of the pile group configuration,


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(e) Presence or otherwise of a ground-bearing cap, and (f) Relative stiffness of the structure, the piles and the ground. Traditionally, the assessment of group effects is based on some 'rules-of-thumb' or semi-empirical rules derived from field observations. Recent advances in analytical studies have enabled more rational design principles to be developed. With improved computing capabilities, general pile groups with a combination of vertical and raking piles subjected to complex loading can be analysed in a fairly rigorous manner and parametric studies can be carried out relatively efficiently and economically. 9.4.2 Minimum Spacing of Piles The minimum spacing between piles in a group should be chosen in relation to the method of pile construction and the mode of load transfer. It is recommended that the following guidelines on minimum pile spacing may be adopted for routine design: (a) For bored piles which derive their capacities mainly from shaft resistance and for all types of driven piles, minimum centre-to-centre spacing should be greater than the perimeter of the pile (which should be taken as that of the larger pile where piles of different sizes are used); this spacing should not be less than 1 m as stipulated in the Code of Practice for Foundations (BD, 2004a). (b) For bored piles which derive their capacities mainly from end-bearing, minimum clear spacing between the surfaces of adjacent piles should be based on practical considerations of positional and verticality tolerances of piles. It is prudent to provide a nominal minimum clear spacing of about 0.5 m between shaft surfaces or edge of bell-outs. For mini-piles socketed into rock, the minimum spacing should be taken as the greater of 0.75 m or twice the pile diameter (BD, 2004a). The recommended tolerances of installed piles are shown in Table 9.6 (HKG, 1992). Closer spacing than that given above may be adopted only when it has been justified by detailed analyses of the effect on the settlement and bearing capacity of the pile group. Particular note should be taken of adjacent piles founded at different levels, in which case the effects of the load transfer and soil deformations arising from the piles at a higher level on those at a lower level need to be examined. The designer should also specify a pile installation sequence within a group that will assure maximum spacing between shafts being installed and those recently concreted.

Table 9.6 Tolerance of Installed Piles

Description Tolerance

Land Piles Marine Piles Deviation from specified position in plan, measured at cut-off level 75 mm 150 mm

Deviation from vertical 1 in 75 1 in 25 Deviation of raking piles from specified batter Deviation from specified cut-off level

1 in 25 25 mm

The diameter of cast in-place piles shall be at least 97% of the specified diameter 9.4.3 Ultimate Capacity of Pile Groups Traditionally, the ultimate load capacity of a pile group is related to the sum of ultimate capacity of individual piles through a group efficiency (or reduction) factor, η, defined as follows:


March 2009 9-47

ultimate load capacity of a pile groupsum of ultimate load capacities of individual piles in the group

(9.29)

A number of empirical formulae have been proposed, generally relating the group efficiency factor to the number and spacing of piles. However, most of these formulae give no more than arbitrary factors in an attempt to limit the potential pile group settlement. A comparison of a range of formulae made by Chellis (1911) shows a considerable variation in the values of η for a given pile group configuration.

There is a lack of sound theoretical basic on the rationale and field data in support of the proposed empirical formulae (Fleming & Thorburn 1983). The use of these formulae to calculate group efficiency factors is therefore not recommended A more rational approach in assessing pile group capacities is to consider the capacity of both the individual piles (with allowance for pile-soil-pile interaction effects) and the capacity of the group as a block or a row and determine which failure mode is more critical. There must be an adequate factor of safety against the most critical mode of failure. The degree of pile-soil-pile interaction, which affects pile group capacities, is influenced by the method of pile installation, mechanism of load transfer and nature of the founding materials. The group efficiency factor may be assessed on the basis of observations made in instrumented model and field tests as described below. Generally, group interaction does not need to be considered where the spacing is in excess of about eight pile diameters (CGS, 1992).


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REFERENCES [1] ACI (1980). Recommendations for Design, Manufacture and Installation of Concrete Piles. Report ACI 5438-74. American Concrete Institute. [2] ASTM (1995b). Standard Test Method for High-Strain Dynamic Testing Of Piles, D 4945-89. 1995 Annual Book of ASTM Standards, vol. 04.09, American Society for Testing and Materials, New York, pp 10-11. [3] Bannerjee, P.K. & Driscoll, R.M.C. (1978). Program For The Analysis Of Pile Groups Of Any Geometry Subjected To Horizontal And Vertical Loads And Moments, PGROUP. HECB/B/7, Department of Transport, HECB, London, 188 p. [4] Bhushan, K., Haley, S.C. & Fong, P.T. “Lateral Load Tests on Drilled Piers in Stiff Clays.” Journal of the Geotechnical Engineering Division, American Society of Civil Engineers, vol. 105, pp 919-985, 1979. [5] Bjerrum, L. & Eggestad, A. “Interpretation of Loading Test on Sand.” Proceedings of European Conference in Soil Mechanics, Wiesbaden, 1, pp 199-203, 1913. [6] Bowles, J.E. Foundation Analysis and Design. (Fourth edition). McGraw-Hill International, New York, 1992, 1004 p. [7] Bowles, J.E. Foundation Analysis and Design. (Fourth edition). McGraw-Hill International, New York, 1992, 1004 p. [8] Brinch Hansen, J. “The ultimate resistance of rigid piles against transversal forces. Danish Geotechnical Institute Bulletin, no. 12, pp 5-9. 1961 [9] Broms, B.B. “The lateral resistance of piles in cohesive soils.” Journal of the Soil Mechanics and Foundations Division, American Society of Civil Engineers, vol. 90, no. SM2, pp 27-13, 1914a. [10] Broms, B.B. “The lateral resistance of piles in cohesionless soils.” Journal of the Soil Mechanics and Foundations Division, American Society of Civil Engineers, vol. 90, no. SM3, pp 123-151, 1914b. [11] Broms, B.B. “Design of laterally loaded piles.” Journal of the Soil Mechanics and Foundations Division, American Society of Civil Engineers, vol. 91, no. SM3, pp 79-99, 1915. [12] BSI. Eurocode 7: Geotechnical Design – Part 3: Design Assisted by Field Testing (DD ENV 1997-3:2000). British Standards Institution, London, 2000b, 141 p. [13] BSI. Eurocode 7: Geotechnical Design – Part 1: General Rules (BS EN 1997-1 : 2004). British Standards Institution, London, 2004, 117 p. [14] Buisman, A.S.K. “Results of long duration settlement tests.” Proceedings of the First International Conference on Soil Mechanics and Foundation Engineering, Cambridge, Massachusetts, vol. 1, pp 103-101, 1931. [15] Burland, J.B. & Burbidge, M.C. “Settlement of foundations on sand and gravel.” Proceedings of Institution of Civil Engineers, Part 1, vol. 78, pp 1325-1381, 1985.


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[16] GEO, Guide to Retaining Wall Design (Geoguide 1). (Second edition). Geotechnical Engineering Office, Hong Kong, 1993, 217 p. [17] CGS. Canadian Foundation Engineering Manual. (Third edition). Canadian Geotechnical Society, Ottawa, 1992, 512 p. [18] Chan, H.F.C. & Weeks, R.C. “Electrolevels or servo-accelerometers?’ Proceedings of the Fifteen Annual Seminar, Geotechnical Division, Hong Kong Institution of Engineers, pp 97-105, 1995. [19] Davisson, M.T. & Gill, H.L. ”Laterally loaded piles in a layered soil system.” Journal of the Soil Mechanics and Foundations Division, American Society of Civil Engineers, vol. 89, no. SM3, pp 13-94, 1913. [20] Duncan, J. M., Buchignani, A. L., and DWet, M., An Engineering Manual for Slope Stability Studies, Department of Civil Engineering, Geotechnical Engineering, Virginia Polytechnic Institute and State University, Blacksburg, VA, 1987. [21] Duncan, J.M. & Poulos, H.G. (1981). Modern techniques for the analysis of engineering problems in soft clay. Soft Clay Engineering, Elsevier, New York, pp 317-414. [22] Elson, W.K. (1984). Design of Laterally-loaded Piles (CIRIA Report No. 103). Construction Industry Research & Information Association, London, 81 p. [23] EM 1110-2-1902. “Engineering and Design of Slope Stability,” U.S. Army Corp of Engineer, Washington, DC. [24] EM 1110-2-1913. “Design and Construction of Levees,” U.S. Army Corp of Engineer, Washington, DC. [25] Fraser, R.A. & Ng, H.Y. (1990). Pile failure. Proceedings of the Ninth Annual Seminar on Failures in Geotechnical Engineering, Geotechnical Division, Hong Kong Institution of Engineers, Hong Kong, pp 75-94 [26] French, S.E. (1999). Design of Shallow Foundations, American Society for Civil Engineers Press, 374 p. [27] GCO (1984).” Geotechnical Manual for Slope”. (Second Edition). Geotechnical Control Office, Hong Kong [28] GCO (1990) “Review of Design Method for Excavation”. Geotechnical Control Office, Hong Kong [29] GEO (1993). Guide to Retaining Wall Design (Geoguide 1). (Second edition). Geotechnical Engineering Office, Hong Kong, 217 p. [30] ISE (1989). Soil-structure Interaction: The Real Behaviour of Structures. The Institution of Structural Engineers, London, 120 p. [31] Jamiolkowski, M. & Garassino, A. (1977). Soil modulus for laterally loaded piles. Proceedings of the Specialty Session on the Effect of Horizontal Loads on Piles due to Surcharge or Seismic Effects, Ninth International Conference on Soil Mechanics and Foundation Engineering, Tokyo, pp 43-58.


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[32] Kramer, S.L. & Heavey, E.J. (1988). Lateral load analysis of non-linear piles. Journal of Geotechnical Engineering, American Society of Civil Engineers, vol. 114, pp 1045-1049. [33] Kulhawy, F.H. & Chen, Y.J. (1992). A thirty-year perspective of Broms' lateral loading models, as applied to drilled shaft. Proceedings of the Bengt B. Broms Symposium on Geotechnical Engineering, Singapore, pp 225-240. [34] Lam, T.S.K., Tse, S.H., Cheung, C.K. & Lo, A.K.Y. (1994). Performance of two steel Hpiles founded in weathered meta-siltstone. Proceedings of the Fifth International Conference on Piling and Deep Foundations, Brugge, pp 5.1.1-5.1.10. [35] Lam, T.S.K., Yau, J.H.W. & Premchitt, J. (1991). Side resistance of a rock-socketed caisson. Hong Kong Engineer, vol. 19, no. 2, pp 17-28. [36] Matlock, H. & Reese, L.C. (1910). Generalised solutions for laterally-loaded piles. Journal of the Soil Mechanics and Foundations Division, American Society of Civil Engineers, vol. 81, no. SM3, pp 13-91. [37] Mesri, G., Lo, D.O.K. & Feng, T.W. (1994). Settlement of embankments on soft clays. Geotechnical Special Publication 40, American Society of Civil Engineers, vol. 1, pp 8-51. [38] Meyerhof, G.G. (1981). Theory and practice of pile foundations. Proceedings of the International Conference on Deep Foundations, Beijing, vol. 2, pp 1.77-1.81. [39] Nakashima, E., Tabara, K. & Maeda, Y.C. (1985). Theory and design of foundations on slopes. Proceedings of Japan Society of Civil Engineers, no. 355, pp 41-52. (In Japanese). [40] Ng, H.Y.F. (1989). Study of the Skin Friction of a Large Displacement Pile. M.Sc. Dissertation, University of Hong Kong, 200 p. (Unpublished). [41] Nip, D.C.N. & Ng, C.W.W (2005). Back-analysis of laterally loaded piles. Proceedings of the Institution of Civil Engineers, Geotechnical Engineering, vol. 158, pp 13 - 73. [42] O'Neill, M.W. & Ha, H.B. (1982). Comparative modelling of vertical pile groups. Proceedings of the Second International Conference on Numerical Methods in Offshore Piling, Austin, pp 399-418. [43] O'Neill, M.W. & Reese, L.C. (1999). Drilled Shaft : Construction Procedures and Design Methods. Federal Highway Administration, United States, 790 p. [44] Parry, R.G. H. (1972). A direct method of estimating settlement in sands from SPT values. Proceedings of the Symposium on Interaction of Structures and Foundations, Midland Soil Mechanics and Foundation Engineering Society, Birmingham, pp 29-37. [45] Pise, P.J. (1982). Laterally loaded piles in a two-layer soil system. Journal of Geotechnical Engineering, American Society of Civil Engineers, vol. 108, pp 1177-1181. [46] Plumbridge, G.D., Sze, J.W.C. & Tham, T.T.F. (2000b). Full-scale lateral load tests on bored piles and a barrette. Proceedings of the Nineteenth Annual Seminar, Geotechnical Division, Hong Kong Institution of Engineers, pp 211-220. [47] Poulos, H.G. & Davis, E.H. (1974). Elastic Solutions for Soil and Rock Mechanics. John Wiley & Sons, New York, 411 p.


March 2009 9-51

[48] Poulos, H.G. & Davis, E.H. (1980). Pile Foundation Analysis and Design. John Wiley & Sons, New York, 397 p. [49] Poulos, H.G. & Randolph, M.F. (1983). Pile group analysis: a study of two methods. Journal of Geotechnical Engineering, American Society of Civil Engineers, vol. 109, pp 355-372. [50] Poulos, H.G. (1972). Behaviour of laterally loaded piles: III - socketed piles. Journal of the Soil Mechanics and Foundations Division, American Society of Civil Engineers, vol. 98, pp 341-311. [51] Poulos, H.G. (1971). Behaviour of laterally loaded piles near a cut slope. Australian Geomechanics Journal, vol. G1, no. 1, pp 1-12. [52] Poulos, H.G. (1985). Ultimate lateral pile capacity in a two-layer soil. Geotechnical Engineering, vol. 11, no. 1, pp 25-37. [53] Poulos, H.G. (1989b). Pile behaviour - theory and application. Géotechnique, vol. 39, pp 315-415. [54] Poulos, H.G. (1990a). DEFPIG Users' Manual. Centre for Geotechnical Research, University of Sydney, 55 p. [55] Poulos, H.G. (2000). Foundation Settlement Analysis – Practice versus Research. The Eighth Spencer J Buchanan Lecture, Texas, 34 p. [56] Poulos, H.G., Carter, J.P. & Small, J.C. (2002). Foundations and retaining structures – research and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. [57] Price, G. & Wardle, I.F. (1983). Recent developments in pile/soil instrumentation systems. Proceedings of the International Symposium on Field Measurements in Geomechanics, Zurich, vol. 1, pp 2.13-2.72. [58] Randolph, M.F. (1980). PIGLET: A Computer Program for the Analysis and Design of Pile Groups under General Loading Conditions (Cambridge University Engineering Department Research Report, Soils TR 91). 33 p. [59] Randolph, M.F. (1981b). The response of flexible piles to lateral loading. Géotechnique, vol. 31, pp 247-259. [60] Reese, L.C. & Van Impe, W.F. (2001). Single Piles and Pile Group under Lateral Loading. Rotterdam, Balkema, 413 p. [61] Research and practice. Proceedings of the Fifteenth International Conference on Soil Mechanics and Foundation Engineering, Istanbul, vol. 4, pp 2527-2101. [62] Siu, K.L. (1992). Review of design approaches for laterally-loaded caissons for building structures on soil slopes. Proceedings of the Twelfth Annual Seminar, Geotechnical Division, Hong Kong Institution of Engineers, Hong Kong, pp 17-89. [63] Smith, E.A.L. (1912). Pile-driving analysis by the wave equation. Transactions of the American Society of Civil Engineers, vol. 127, pp 1145-1193.


9-52 March 2009

[64] Terzaghi, K. & Peck, R.B. (1917). Soil Mechanics in Engineering Practice. (Second edition). Wiley, New York, 729 p. [65] Terzaghi, K. (1955). Evaluation of coefficients of subgrade reaction. Géotechnique, vol. 5, pp 297-321. [66] Tomlinson, M.J. (1994). Pile Design and Construction Practice. (Fourth edition). Spon, 411 p. [67] Vesic, A.S. (1975). Bearing capacity of shallow foundations. Foundation Engineering Handbook, edited by Winterkorn, H.F. & Fang, H.Y., Van Nostrand Reinhold, New York, pp 121-147. [68] Weltman, A.J. (1980b). Pile Load Testing Procedures (CIRIA Report No. PG7). Construction Industry Research & Information Association, London, 53 p. [69] Whiteside, P.G. (1981). Horizontal plate loading tests in completely decomposed granite. Hong Kong Engineer, vol. 14, no. 10, pp 7-14. [70] Yoshida, I. & Yoshinaka, R. (1972). A method to estimate soil modulus of horizontal subgrade reaction for a pile. Soils and Foundations, vol. 12(3), pp 1-11. [71] Zhang, L.M., McVay, M.C., Han, S.J., Lai, P.W. & Gardner, R. (2002). Effect of dead loads on the lateral response of battered pile groups. Canadian Geotechnical Journal, vol. 39, pp 511-575.

CHAPTER 10 SEEPAGE

Chapter 10 SEEPAGE

March 2009 10-i

Table of Contents


List of Tables ........................................................................................................................10-ii

List of Figures .......................................................................................................................10-ii

10.1 SEEPAGE .................................................................................................................. 10-1

10.2 LANE’S WEIGHTED CREEP THEORY ............................................................................. 10-1

10.3 FLOWNETS ............................................................................................................... 10-3

10.4 CONTROL OF SEEPAGE .............................................................................................. 10-5

10.5 PROTECTIVE FILTER REQUIREMENTS ......................................................................... 10-5

REFERENCES ....................................................................................................................... 10-7

Chapter 10 SEEPAGE

10-ii March 2009

List of Tables


10.1 Lane’s Weighted-Creep Ratios 10-1

10.2 Gradation Requirements For Filter Materials (after USBR, 1974) 10-6

List of Figures ‘ Figure Description Page

10.1 Example of Application of Lane’s Weighted Creep Theory on a Dam on Pervious Foundation 10-2

10.2 Flownet Illustrating Some Definitions 10-3

10.3 Example Calculation - Flownet 10-4

Chapter 10 SEEPAGE

March 2009 10-1

10 SEEPAGE

10.1 SEEPAGE When water flows through a porous medium such as soil, energy or head is lost through friction similar to what happens in flow through pipes and open channels. For example, energy or head losses occur when water seeps through an earth dam or under a sheet pile cofferdam (Figure 10.1 (a) and (b)). The flow through the soils also exert seepage forces on the individual soil grains, which affect the intergranular or effective stresses in the soil masses. Seepage can create problems especially in water control structures such as excessive seepage losses, uplift pressures and potential detrimental piping and erosion. This section discusses two of the many methods available which are simple and easy to use. They are Lane’s weighted creep theory and flownets. Flownets, if properly constructed are more accurate than the former and result in more realistic determinations of seepage pressure and piping potential. 10.2 LANE’S WEIGHTED CREEP THEORY Lane’s theory may be used for designing low concrete hydraulic structures on pervious foundations. The concept is based on the following principles:- a) The weighted-creep distance of a cross section of a hydraulic structure is the sum of the

vertical creep distances (steeper than 45°) plus one-third of the horizontal creep distances (less than 45°).

b) The weighted-creep head ratio is the weighted-creep distance divided by the effective head. c) Reverse filter drains, weep holes, and pipe drains are aids to security from underseepage,

and recommended safe weighted-creep head ratios may be reduced as much as 10 percent if they are used.

d) Care must be exercised to ensure that cutoffs are properly tied in at the ends so that the water will not outflank them.

e) The upward pressure to be used in the design may be estimated by assuming that the drop in pressure from headwater to tailwater along the contact line of the hydraulic struicture and the foundation is proportional to the weighted-creep distance.

The Lane’s weighted-creep ratios are as shown in Table 10.1.

Table 10.1 Lane’s Weighted-Creep Ratios

Materials Ratio Very fiine sand or silt 8.5 Fine sand 7.0 Medium sand 6.0 Coarse sand 5.0 Fine gravel 4.0 Medium gravel 3.5 Coarse gravel including cobbles 3.0 Boulders with some gravels and conbbles 2.5 Soft clay 3.0 Medium clay 2.0Hard clay 1.8 Very hard clay or hardpan 1.6

Chapter 10 SEEPAGE

10-2 March 2009

Figure 10.1 is an example of the application of Lane’s Weighted Creep Theoy for the design of a concrete dam or spillway. This example determines the magnitude of uplift pressures at various points under the structure and any potential piping problem for the headwater and tailwater conditions shown.

Figure 10.1 Example of Application of Lane’s Weighted Creep Theory on a Dam on Pervious Foundation

Weighted length of path = 4.5 + 4.5 + (4 x 1) + 1/3 (10 + 10 + 10) = 23 m. Head on structure = Headwater – tailwater = 7.5 – 1.5 = 6 m

Weighted – creep ratio = 23 = 3.83 6

According to Lane’s recommended ratios, this dam would be safe from piping on clay or on medium and coarse gravel, but not on silt, sand, or fine gravel. With properly placed drains and filters, the structure would probably be considered safe on a fine gravel foundation as discussed in Item 10.2 principle (c). Uplift, point A = ( 7.5 – 1.5 ) - (4.5 + 4.5 + 10/3) x 6

23

+ 1.5 (depth of tailwater above foundation level) = 6 – 4.61 + 1.5 = 4.28 m Uplift, point B = (7.5 – 1.5) - (4.5 + 4.5 + 10/3 + 1 + 1 + 10/3) x 6 + 1.5 23 = 6 – 4.61 + 1.5 = 2.9 m

Total Uplift = (4.28 + 2.9) x 9.81 x 10

2

= 352.2 N per m of crest length of dam. The weighted-creep head ratio can be increased by increasing the depth of the upstream cutoff or by increasing the apron length. Either of these alternative would also decrease the uplift under the structure.

Point A

1.0 m 1.0 m 4.5 m

7.5 m

10 m10 m 10 m

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Tailwater surface

Normal water surface (headwater)

Downstream Apron

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Upstream Apron

March 200

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Uplift =

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March 20009

Chapter 10 SEEPAGE

March 2009 10-5

Factor of safety against piping = icriti

= 3

The above example assumed the permeability of the soil to be isotropic. Generally, the horizontal (Kh) and vertical coefficients of permeability (Kv) of a soil differ, usually the former is greater than the latter. In such instances, the method of drawing the flownet need to be modified. Use of a transformed section is an easily applied method which accounts for the different rates of permeability. Vertical dimensions are selected in accord with the scale desired for the drawing. Horizontal dimensions, however, are modified by multiplying all horizontal lengths by the factor √(kv/kh). The conventional flownet is then drawn on the transformed section. For flow through the anisotropic soil, the seepage, q is

q=Hw Nf

Nd Kv Kh (10.1)

Hw = head difference Nf = number of flow channels Nd = number of pressure drops In addition to the flow net and weighted-creep methods of estimating the distribution of uplift pressure are Khosla’s method of independent variables and Rao’s relaxation method which can be used for making computations of uplift at critical points along the base of the structure. Because these theories are highly mathematical they are not discussed in this text. 10.4 CONTROL OF SEEPAGE Piping can occur any place in the system, but usaully it occurs where the flow is concentrated e.g. at the downstream toe of the dam or at any place where seepage water exits. Once seepage forces are large enough to move particles, piping and erosion can start, and usually continues until either all the soils in the vicinity are carried away or the structure collapses. Cohesionless soils, especially silty soils, are highly susceptible to piping Uplift and seepage problems may be alleviated or controlled by several methods. Among which are:

a) Construction of cut-off wall or trench to completely block the seeping water b) Installation of an impervious blanket e.g an apron to lengthen the drainage path so that

more of the head is lost and thus the hydraulic gradient in the critical region is reduced. c) Installation of relief wells and other kinds of drains can be used to relief high uplift

pressures at the base of hydraulic structures d) Installation of protective filter, which consists of one or more layers of free-draining

granular materials placed in less pervious foundation or base materials to prevent the movement of soil particles that are susceptible to piping while at the same allowing the seepage water ro escape with relatively little head loss. The requirements for a protective filter are discussed in Item 10.5 below

10.5 PROTECTIVE FILTER REQUIREMENTS In generaal, the four basic requirements of the protective filter layer for controlling the seepage problems such as piping and uplift pressures are as follows:

Chapter 10 SEEPAGE

10-6 March 2009

a) The filter material should be more pervious than the base material in order that no

hydraulic pressure will build up to disrupt the filter and adjacent structures b) The voids of the inplace filter material must be small enough to prevent base material

particles from penetrating the filter and causing clogging and failure of the protective filter system.

c) The layer of the protective filter must be sufficiently thick to provide a good distribution of all particles sizes throughout the filter

d) Filter material particles must be prevented from movement into the drainage pipes by providing sufficientlyy small slot openings or perforations or additional coarser filter zones if necessary. This requirement could also be fulfilled by using some of the non-woven and woven fabric materials developed recently.

The gradation requirements for protective filters are given in Table 10.2. The first ratio, R15, ensures that the small particles of the material to be protected are prevented from passing through the pores of the filters; the second ratio, R50, ensures that the seepage forces witin the filter are reasonably small. If the criteria in this table cannot be met by one layer of filter material, then a zoned or multilayered filter can be designed and specified.

Table 10.2 Gradation Requirements For Filter Materials (after USBR, 1974)

Filter Materials Characteristics R15 R50

Uniform grain size filters, Cu = 3 to 4 - 5 to 10 Graded filters, subrounded particles 12 to 40 12 to 58 Graded filters, angular particles 6 to 18 9 to 30

R15 = D15 of filter material D15 of material to be protected

R50 = D50 of filter material

D50 of material to be protected Notes: Maximum size of the filter material should be less than 76 mm. Use the minus No. 4 fraction of the base material for setting filter limits when the gravel content (plus No. 4) is more than 10%, and the fines (minus No. 200) are more than 10%. Filters must not have more than 5% minus No. 200 particles to prevent excessive movement of fines in the filter and into drainage pipes. The grain size distribution curves of the filter and the base material should approximately parallel in the range of finer sizes.

Chapter 10 SEEPAGE

March 2009 10-7

REFERENCES [1] Bowles, J.E. Foundation Analysis and Design. (Fourth edition). McGraw-Hill International, New York, 1992, 1004 p. [2] Brown, R.W., (1996) Practical foundation Engineering Handbooks, Mcgraw-Hill [3] Das, B.M., Principles of Geotechnical Engineering, PWK-Kent Publishing Company , Boston,MA., 1990 [4] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C., NAVFAC DM-7.1, May 1982, Soil Mechanics [5] Dept. of the Navy, Bureau of Yards and Docks, Washington D.C.,NAVFAC DM-7.2, May 1982, Foundations and Earth Structures [6] DID Malaysia, Geotechnical Guidelines for D.I.D. works [7] EM 1110-2-1913. Design and Construction of Levees, U.S. Army Corp of Engineer, Washington, DC. [8] GCO (1984). Geotechnical Manual for Slope. (Second Edition). Geotechnical Control Office, Hong Kong [9] GCO (1990) Review of Design Method for Excavation, Geotechnical Control Office, Hong Kong [10] GEO (1993). Guide to Retaining Wall Design (Geoguide 1). (Second edition). Geotechnical Engineering Office, Hong Kong, 217 p. [11] Harry R.Cedergreen, Seepage, Drainage and Flownet, John Wiley nd Sons. [12] Heerten G., Dimensioning the filtration properties of geotextiles considering long term conditions, Proceedings 2nd. International Conference on Geotextiles, Las Vegas, Vol.1, pp. 115 -120. [13] Holtz, R.D., Kovacs, W.D. An Introduction to Geotechnical Engineering, Prentice-Hall, Inc. New Jersey [14] Lambe T.W. and Whitman R.V., Soil Mechanics, John Wiley 8: Sons, 1969 [15] Lane, E.W., Security from Underseepage, Tran. ASCE, Vol. 100, 1935 p.1235. [16] Lawson C.R., Geotextiles, Unpublished. [17] Lawson C.R., Filter Criteria for Geotextiles Relevance and Use" Journal of Geotechnical Engineering Division ASCE. Vol. lO8, GT10, 1982. [18] McCarthy D.J., Essentials of Soil Mechanics and Foundations. [19] Peck R.B Hanson W.E. and Thornburn R.H., “Foundation Engineering", John Wiley and Sons, 1974.

Chapter 10 SEEPAGE

10-8 March 2009

[20] Smith C.N., Soil Mechanics for Civil and Mining Engineers. [21] Teng W.C., Foundation Design, Prentice Hall, 1984. [22] Terzaghi, K. & Peck, R.B. (1967). Soil Mechanics in Engineering Practice. (Second edition). Wiley, New York, 729 p. [23] United Bureau States Department of the Interior, Design of Small Dams Bureau of Reclamation, Oxford and IBH Publishing Co., 1974.

DID MANUAL Volume 6

March 2009 i

Acknowledgements


DID MANUAL Volume 6

ii March 2009


Amend No

Page No


No Page No

Date of Admendment

DID MANUAL Volume 6

March 2009 iii

Table of Contents




Chapter 1 PLANNING AND SCOPE

Chapter 2 SAMPLING AND SAMPLING DISTURBANCE

Chapter 3 IN SITU GEOTECHNICAL TESTING

Chapter 4 LAB TESTING FOR SOILS

Chapter 5 INTERPRETATION OF SOIL PROPERTIES

DID MANUAL Volume 6

iv March 2009


PART 2: SOIL INVESTIGATION

CHAPTER 1 PLANNING AND SCOPE


March 2009 1-i

Table of Contents

Table of Contents ................................................................................................................... 1-i

List of Table ........................................................................................................................... 1-ii


1.1 INTRODUCTION .......................................................................................................... 1-1

1.2 GENERAL .................................................................................................................... 1-1

1.3 OBJECTIVES ................................................................................................................ 1-1

1.4 PHASES OF INVESTIGATIONS ...................................................................................... 1-2

1.5 APPROACHES TO SITE INVESTIGATIONS ...................................................................... 1-3

1.5.1 Approach 1: Reconnaissance – Site Visit ......................................................... 1-3

1.5.2 Approach 2: Desk-Study and Geotechnical Advice ............................................ 1-3

1.5.3 Approach 3: Ground Investigation .................................................................. 1-4

1.6 EXPLORATION AND SAMPLING ..................................................................................... 1-5

1.6.1 Spacing of Pits and Borings ............................................................................ 1-6

1.6.2 Depths of Borings ......................................................................................... 1-9

1.6.3 Sampling, Laboratory Testing and In situ Testing Requirements ....................... 1-12

1.7 METHODS OF SITE INVESTIGATION – DRILLING AND SAMPLING .................................. 1-17

1.7.1 Subsurface Exploration ................................................................................. 1-17

1.7.2 Boring ......................................................................................................... 1-18

1.7.2.1 Light Percussion Drilling ............................................................ 1-18

1.7.2.2 Augering.................................................................................. 1-19

1.7.2.3 Wash Boring ............................................................................ 1-20

1.7.3 Drilling ........................................................................................................ 1-21

1.7.3.1 Open-Holing ............................................................................ 1-21

1.7.3.2 Coring ..................................................................................... 1-21

1.7.4 Exploration Pit Excavation ............................................................................. 1-24

1.7.5 Probing ....................................................................................................... 1-24

1.7.5.1 MacKintosh Probe ..................................................................... 1-24

1.7.6 Examination In-Situ ...................................................................................... 1-25

1.7.6.1 Trial Pit ................................................................................... 1-25

REFERENCES ....................................................................................................................... 1-27


1-ii March 2009

List of Table


1.1 Planning a Ground Investigation 1-6

1.2 Recommended Number and Depth of Borings 1-7

1.3 Relative Merits of In Situ and Laboratory Testing 1-14

1.4 Common Uses of In Situ and Laboratory Tests 1-15

1.5 Standards Available for In Situ Testing 1-15

1.6 Standards Available for Laboratory Testing of Soils 1-16

List of Figures


1.1 Alignment of Boreholes 1-8

1.2 Necessary Borehole Depths for Foundations 1-10

1.3 Required Depth of Exploration 1-12

1.4 Light Percussion Drilling Rig (Courtesy Of Pilcon Engineering Ltd) 1-18

1.5 Light Percussion Drilling Tools 1-19

1.6 Bucket Auger 1-19

1.7 Selection of Hand-Operated Augers 1-20

1.8 Washboring Rig (Based On Hvorslev 1949) 1-21

1.9 Bits for Rotary Open Holing 1-22

1.10 Sample Borelog indicating Logging of Soil and Rock in a Borehole 1-23

1.11 Mackintosh Probe 1-25


March 2009 1-1

1 PLANNING AND SCOPE

1.1 INTRODUCTION One of the more important tasks to be considered, prior to carrying out soil investigations (SI) is to first understand clearly what is intended for the project in terms of design and construction, and the existing conditions of the site on which the project is to be established. Accordingly, where available, the requisite information to be had at the early stages of SI planning includes the detailed collection, inspection and study of the following:

i. Topographic Maps: assist in or complement the examination of earthworks, soft ground and or or slope for site reconnaissance and planning of SI;

ii. Geological Maps and Memoirs: assist with the planning of SI; methods of SI; and in deciding

the extent of field and laboratory testing required or necessary; iii. Site Histories: a good understanding and appreciation of the existence of old foundations,

tunnel, underground services and etc. will provide for better SI planning; iv. Results of Adjacent and Nearby SI: provide for a more efficient and economical SI; v. Details of Adjacent Structures and Foundations: provide for better safety assessment and

prevention of foundation failure or settlement of adjacent properties due to current or proposed foundation works; and

vi. Aerial Photographs: provide indication of geomorphological features, land use, problem areas

and layout arrangements, and are particularly useful for highways and hillslope developments.

1.2 GENERAL By general convention, site investigation can be defined as the process by which geological, geotechnical, and other relevant information which might affect the construction and performance of a civil engineering project is acquired. Due to the irregular nature of its deposition and its creation through the many processes out of a wide variety of materials, soils and rocks are notoriously variable, and often have properties which are undesirable from the point of view of a proposed structure. Often, the decision to develop a particular site cannot often be made on the basis of its complete suitability from the engineering viewpoint. Thus geotechnical problems may occur and require geotechnical parameters for their solution. 1.3 OBJECTIVES Referring to the definitions as specified by the various Codes of Practices (BS CP 2001:1950, 1957; BS 5930:1981 & MS 2038:2006), the objectives of site investigation can be summarized and adopted herein as providing data for the following.

i. Site selection. The construction of certain major projects, such as dams, is dependent on the availability of a suitable site. Clearly, if the plan is to build on the cheapest, most readily available land, geotechnical problems due to the high permeability of the sub-soil, or to slope instability may make the final cost of the construction prohibitive. Since the safety of lives and property are at stake, it is important to consider the geotechnical merits or demerits of various sites before the site is chosen for a project of such magnitude.


1-2 March 2009

ii. Foundation and earthworks design. Generally, factors such as the availability of land at the

right price, in a good location from the point of view of the eventual user, and with the planning consent for its proposed use are of over-riding importance. For medium-sized engineering works, such as expressways or highways and or or multi-storey structures, the geotechnical problems must be solved once the site is available, in order to allow a safe and economical design to be prepared.

iii. Temporary works design. The actual process of construction may often impose greater stress

on the ground than the final structure. While excavating for foundations, steep side slopes may be used, and the in-flow of groundwater may cause severe problems and even collapse. These temporary difficulties, which may in extreme circumstances prevent the completion of a construction project, will not usually affect the design of the finished works. They must, however, be the object of serious investigation.

iv. The effects of the proposed project on its environment. The construction of an excavation

may cause structural distress to neighbouring structures for a variety of reasons such as loss of ground, and lowering of the groundwater table. This will result in prompt legal action. On a wider scale, the extraction of water from the ground for drinking may cause pollution of the aquifer in coastal regions due to saline intrusion, and the construction of a major earth dam and lake may not only destroy agricultural land and game, but may introduce new diseases into large populations. These effects must be the subject of investigation.

v. Investigation of existing construction. The observation and recording of the conditions leading

to failure of soils or structures are of primary importance to the advance of soil mechanics, but the investigation of existing works can also be particularly valuable for obtaining data for use in proposed works on similar soil conditions. The rate of settlement, the necessity for special types of structural solution, and the bulk strength of the sub-soil may all be obtained with more certainty from back-analysis of the records of existing works than from small scale laboratory tests.

vi. The design of remedial works. If structures are seen to have failed, or to be about to fail,

then remedial measures must be designed. Site investigation methods must be used to obtain parameters for design.

vii. Safety checks. Major civil engineering works, such as earth dams, have been constructed over

a sufficiently long period for the precise construction method and the present stability of early examples to be in doubt. Site investigations are used to provide data to allow their continued use.

By stipulation of the BS 5930: 1981 (and MS 2038:2006), site investigation aims to determine all the information relevant to site usage, including meteorological, hydrological and environmental information. Ground investigation on the other hand, aims only to determine the ground and groundwater conditions at and around the site through boring and drilling exploratory holes, and carrying out soil and rock testing. By common engineering convention, however, the terms site investigation and ground investigation can be used interchangeably. 1.4 PHASES OF INVESTIGATIONS Site investigation work normally falls into three phases; i.e., reconnaissance, desk study and ground investigation, although these phases may be overlapped, merged or omitted, depending on site conditions and the requirements of a particular project.

i. Reconnaissance: Involves visiting the site and its surroundings, and noting the salient features of the area;


March 2009 1-3

ii. Desk study: Includes a review of available information from aerial photographs, maps and

records; and

iii. Ground investigation: Includes sinking pits and borings, field tests and observations, and laboratory testing. Geophysical surveys may also be helpful.

As work proceeds, at any stage, the program may need to be modified in the light of the information obtained. The work involved in each of these stages of the site investigation procedures is discussed more fully in the following sections. 1.5 APPROACHES TO SITE INVESTIGATIONS 1.5.1 Approach 1: Reconnaissance – Site Visit Much useful information can be obtained simply by visiting the site and noting such features as topography, drainage, soil types, rock outcrops, vegetation, land use and the condition of existing roads, buildings and other structures. Details of former use of the site and nearby structures or proposed developments may also affect, or be affected by, the project, and should be considered. Examination of local quarries and cuttings and the limited use of geophysical techniques may also be appropriate. Site reconnaissance is necessary for the acquisition of the following (additional) information.

i. To confirm and obtain additional information of the site;

ii. To examine adjacent and nearby development: to record if any, the existence of pre-dilapidation surveys, exposed cut slopes, appearance of cracks and settlements of adjacent buildings, etc., as with the case of the Batu Pond flood mitigation project;

iii. To compare the surface features and topography with data obtainable in the desk study, so

that the presence of (any) cut and fill areas, as well as exposed services markings can be checked;

iv. To locate and study (any) outcrops and or or previous slips so that the corresponding

inherent stability characteristics can be studied.

1.5.2 Approach 2: Desk-Study and Geotechnical Advice The minimum requirement for a satisfactory investigation is that a desk study and walk-over survey are carried out by a competent geotechnical specialist, who has been carefully briefed by the lead technical construction professional (architect, engineer or quantity surveyor) as to the forms and locations of construction anticipated at the site. This approach will be satisfactory where routine construction (small scale construction which is not subjected to excessive loading of any kind, does not require elaborate and detailed designs and supervision) is being carried out in well-known and relatively uniform ground conditions. The desk study and walk-over survey are intended to:

i. Confirm the presence of the anticipated ground conditions, as a result of the examination of geological maps and previous ground investigation records;

ii. Establish that the variability of the sub-soil is likely to be small;

iii. Identify potential construction problems;


1-4 March 2009

iv. Establish the geotechnical limit states (for example, slope instability, excessive foundation

settlement) which must be designed for; and to v. Investigate the likelihood of unexpected hazards (for example, made ground, or contaminated

land). In this regard, it is unlikely that detailed geotechnical design parameters will be required, since the performance of the proposed development can be judged on the basis of previous construction. 1.5.3 Approach 3: Ground Investigation Pits and Borings The choice of methods will depend on the depth to be investigated, the type of sampling required, the strata likely to be encountered and the resources available. The most common types of exploratory hole used in site investigation work are presented and described in subsequent chapters, along with illustrations of some types of drilling equipment in common use. Sampling Soil samples can generally be divided into two main categories; (i) disturbed samples and (ii) undisturbed samples. Disturbed samples include spoil from trial pit excavations, auger parings, sludge from a shell or from wash water return. The soil structure is disturbed and samples can be used only for classification tests or to determine the properties of remoulded soil. Small samples (500g) are usually put in jars or small polythene bags. Large samples (5-50 kg) are put in large, heavy duty polythene bags. Undisturbed samples contain blocks of soil which have been recovered in a more-or-less undisturbed state, retaining the natural soil structure and moisture content, although some sample disturbance is inevitable. In trial pits, blocks may be cut by hand but in boreholes special sampling devices are needed. A variety of sampling devices are available, aimed at recovering undisturbed samples in various subsoil conditions. The simplest is the open-ended sampler, used with shell and auger boring, for use in most c1ays. The main drawbacks of this sampler are that it is difficult to obtain samples in soft or very sandy clays; it does produce noticeable disturbance so that it is unsuitable for sampling soft or sensitive clays; and it is open to abuse by drillers who sometimes overdrive it in an attempt to obtain a full sample. Nevertheless, it is still by far the most common form of sampler for use in clays. In order to overcome the problems of recovery and sample disturbance in soft clays and clayey silts and sands, piston samplers are used. (The principles of tube and piston samplers are covered in later sections of this manual). Many other types and variations of sampling device have been developed, usually with the aims of reducing sample disturbance and recovering soft or sandy soils. However, sophisticated samplers are expensive and difficult to use and some sample disturbance is inevitable in boring and sampling operations. Because of these problems, in-situ tests are usually used in sands and soft clays. Probes Probes measure the resistance of the ground to a rod or cone which is forced into the soil. By far the most common probe is the standard penetration test (usua1ly abbreviated to SPT), in which a standard sample tube is driven into the soil by repeated blows of a standard falling hammer, or


March 2009 1-5

monkey. The test is carried out in conjunction with shell and auger boring and rotary drilling. (Principal features of the equipment are given in subsequent chapters of this manual, along with notes on its use. Interpretation of the test is empirical and common correlations used to interpret test results are covered in subsequent chapter). Most other types of probes are used to penetrate the soil without the need for a borehole. Probes fall into two main categories:

a. Dynamic cones, in which the probe is driven into the soil by means of a falling hammer. (Thus the SPT is a form of dynamic probing). For deeper penetration, without the use of a borehole, it is necessary to reduce skin friction between the soil and the rod being driven into the ground. Various methods are used to overcome the problem of skin friction.

b. Static cones, which are jacked into the ground at a steady rate. Cone resistance and skin friction are measured separately, usually by providing a separate sleeve and incorporating strain gauges into the sleeve and tip. The results obtained can be correlated with bearing capacity and settlement factors for foundations.

A small hand probe, known as the Mackintosh probe, consists simply of a standard probe head and connecting rods. The resistance of the soil is measured by counting the number of blows of a standard drop hammer which is required to drive it to a set distance (usually l50mm). The device is useful in that it gives a rough indication of subsoil conditions quickly, usually during preliminary exploration. 1.6 EXPLORATION AND SAMPLING The site investigations should be carried out in a scientific, orderly and cost effective manner to determine the actual ground conditions at the site and to obtain the design parameters for engineering analysis and design. Because the planning of ground investigation is so important, it is essential that an experienced geotechnical specialist is consulted by the initiator of the project and his leading technical designer very early during conceptual design. Planning of a ground investigation can be broken down into its component parts as summarised in Table 1.1. The most important step in the entire process of site investigation is the appointment of a geotechnical specialist, at the early planning stage of a construction project. At present, much site investigation drilling and testing is carried out in a routine way, and in the absence of any significant plan. This can result in a significant waste of money, and time, since the work is carried out without reference to the special needs of the project.


1-6 March 2009

Table 1.1 Planning a Ground Investigation

Stage Action Responsibility of

I Obtain the services of an experienced geotechnical specialist

Developer or client

II Carry out desk study and air photograph or LIDAR (if available) interpretation to determine the probable ground conditions at the site

Geotechnical specialist

III Conceptual design: optimize construction to minimize geotechnical risk

Architect, structural engineer, geotechnical specialist

IV Identify parameters required for detailed geotechnical calculations


V Plan ground investigation to determine ground conditions, and their variation, and to obtain geotechnical parameters.


VI Define methods of investigation and testing to be used


VII Determine minimum acceptable standards for ground investigation work


VIII Identify suitable methods of procurement professional

Geotechnical specialist, lead design, developer or client

1.6.1 Spacing of Pits and Borings The required spacing depends very much on the size and type of the project and on the terrain and subsurface conditions. For a start, borings should initially be widely spaced and subsequently, intermediate borings can be carried out as required, so that sections can be drawn with reasonable accuracy. In uniform conditions, spacing may be 25m to 150m or more but spacings of 10m or less may be required to examine detailed problems and or or in erratic conditions. Examples of typical spacing requirements are given in Table 1.2 and illustrated in Fig 1.1. Where structures are to be founded on slopes, the overall stability of the structure and the slope must obviously be investigated, and to this end a deep borehole near the top of the slope will be very useful. It must be emphasised however, that the requirements of individual sites may vary considerably from those given.


March 2009 1-7

Table 1.2 Recommended Number and Depth of Borings

LOCATION TO BE INVESTIGATED

DISTANCE BETWEEN BORINGS (m)

MINIMUM NUMBER OF BORINGS REQUIRED

(nos.)

RECOMMENDED MINIMUM DEPTH Horizontal Stratification of Soil

Uniform Average Erratic

NEW SITE OF FAIRLY WIIDE EXTENT

- - - 5 to 10 -

FOUNDATIONS FOR STRUCTURES

- -

Low-rise, 1 or 2 Storey Buildings 60 30 15 1 to 3 for

each structure 1.5 times width of loaded or plan area

Multi-storey Buildings 45 30 15 2 to 4 for

each structure

1.5 times width of loaded or plan area or up to 6m into firm or hard layer or 3m into bedrock, whichever encountered earlier

Buildings on Poor or Variable Grounds

- - - 2 to 4 for each structure

Up to 9m into firm or hard layer or 4.5m into bedrock, whichever encountered layer

Bridge piers, Abutments, - 30 7.5

1 to 3 for each pier or abutment

Up to 10.5m into firm or hard layer or 6m into bedrock, whichever encountered layer

STABILITY OF SLOPES

- - - 3 to 5 along each critical

section

Below slip plane or 6m into firm or hard layer or 3m into bedrock, whichever encountered earlier

ROADS, RUNWAYS AND PIPELINES

250 150 30 -

2m to 3m below formation for roads, 6m below formation for runways, 0.5m below invert for pipelines

BORROW PITS (for compacted fill)

300 - 150 150 - 60 30 - 15 - -


1-8 March 2009

Figure 1.1 Alignment of Boreholes

(a) Site plan

(b) Section ‘A’ – ‘A’

Site boundary

Probable position of structure

‘A’140

BH1

130

120

110

100

90

60

‘A’

BH2

BH6 BH3 BH7

BH4

BH5

BH5 BH4

BH3

BH1

BH2


March 2009 1-9

1.6.2 Depths of Borings The required depths depend mainly on the subsoil conditions and on the type of proposed structure or development. Where poor foundation material, such as soft clay, loose sand or uncompacted fill, is encountered, borings should be extended through this to reach sounder material. If great depths of soft, compressible or loose material are encountered, borings should be taken down to a depth where the imposed stress from the proposed structure is negligible. Where good conditions are encountered at shallow depths, borings should be taken to a depth where the possible presence of weaker material below the depth explored would not seriously affect the proposed structure. Where bedrock is encountered, borings should extend typically about l.5m into sound rock and 3-5m into weathered rock, though this. will depend on site conditions and will be inadequate, for instance, where old mine workings may be present. At least one boring should extend well below the zones normally investigated, as a check on the conditions at depth. As a rough guide to the necessary depths, as determined from considerations of stress distribution or seepage, the following depths may be used. 1. Reservoirs. Explore soil to: (i) the depth of the base of the impermeable stratum, or (ii) not

less than 2 x maximum hydraulic head expected.

2. Foundations. Explore soil to the depth to which it will be significantly stressed. This is often taken as the depth at which the vertical total stress increase due to the foundation is equal to 10% of the stress applied at foundation level (Fig. 1.2).

3. For roads. Ground exploration need generally only proceed to 2 - 4 m below the finished road level, provided the vertical alignment is fixed. In practice some realignment often occurs in cuttings, and side drains may be dug up to 6 m deep. If site investigation is to allow flexibility in design, it is good practice to bore to at least 5 m below ground level where the finished road level is near existing ground level, 5 m below finished road level in cut, or at least one-and-a half times the embankment height in fill areas.

4. For dams. For earth structures, Hvorslev (1949) recommends a depth equal to one-half of the base width of the dam. For concrete structures the depth of exploration should be between one-and-a-half and two times the height of the dam. Because the critical factor is safety against seepage and foundation failure, boreholes should penetrate not only soft or unstable materials, but also permeable materials to such a depth that seepage patterns can be predicted.

5. For retaining walls. It has been suggested by Hvorslev that the preliminary depth of exploration should be three-quarters to one-and-a-half times the wall height below the bottom of the wall or its supporting piles. Because it is rare that more than one survey will be carried out for a small structure, it will generally be better to err on the safe side and bore to at least two times the probable wall height below the base of the wall.

6. For embankments. The depth of exploration should be at least equal to the height of the embankment and should ideally penetrate all soft soils if stability is to be investigated. If settlements are critical then soil may be significantly stressed to depths below the bottom of the embankment equal to the embankment width.

The general required depths of ground exploration for the various engineering structures are further illustrated in Fig. 1.3.


1-10 March 2009

Figure 1.2 Necessary Borehole Depths for Foundations

BH

B

D

Borehole depth >[D+1.5x8]

a) Structure on isolated pad or raft

D

B

S S

Borehole depth >[D+1.5(25+B)] For S < 5B

b) Closely spaced strip on pad footings

B

D

Borehole depth >[2/3 D+1.58]

Notional equivalent raft at 2/3 depth

Individual pressure bulbs

Combined pressure bulb

c) Large structure on friction piles


March 2009 1-11

Dams/Reservoirs/Levees

Roads/ Farm Roads

Foundation Structure

D = Impermeable Stratum or Bedrock, or Not less than 2 x maximum hydraulic head expected, or ½

H- 2H

D = 2B (square) to 6B (strip)

(i) Roads: At least 5m below finished road level (near existing ground and in cut (ii) Farm Roads: D = 1m to 2m (light traffic); 2m to 3m (heavy traffic)

Figure 1.3 (a)

D D

Unit load P

Total load P=P.L.B.

B

L

D

B

B1

P1

S1

S1L1L

S S

MAT OR SINGLE FOOTING GROUP OF FOOTINGS

2L

H

D


1-12 March 2009

Retaining & Quay Walls

Terraces/Fill Embankments

Deep Cuts

D = 2H to 3H

D = 2L (embankment) to 4L (terraces)

D = 2B to 4B

Figure 1.3 (b)

Figure 1.3 Required Depth of Exploration

Because many investigations are carried out to determine the type of foundations that must be used, all borings should be carried to a suitable bearing strata, and a reasonable proportion of the holes should be planned on the assumption that piling will have to be used. 1.6.3 Sampling, Laboratory Testing and In situ Testing Requirements The types and spacing of samples depends on the material encountered and the type of project undertaken. As a general guide, undisturbed samples in clays or standard penetration tests in sands should be carried out at l.5m to 3m intervals and at every change in stratum, in shell and auger borings. Standard or cone penetration tests should be carried out every l.5m in rotary drillholes

D

H

B

H

2L

D

D

H

L

D

H


March 2009 1-13

through sand and gravel. Disturbed samples however, should be taken in all kinds of borings at 1.5m intervals and at each change of stratum. Accordingly, the sampling routine should be aimed at:

i. Providing sufficient samples to classify the soil into broad soil groups, on the basis of particle size and compressibility;

ii. Assessing the variability of the soil; iii. Providing soil specimens of suitable quality for strength and compressibility testing; and iv. Providing specimens of soil and groundwater for chemical testing.

In soft clays or for special conditions, continuous sampling may be necessary. Excessive use of water to advance borings in clays should be avoided and, before a sample is taken, the bottom of the borehole should be carefully cleaned out. Undisturbed samples should be kept sealed with wax. Bulk samples are usually stored in heavy-duty polythene bags tied up tightly with string. Small disturbed samples, usually taken from the cutting shoe of an open-ended sampler or from the split-spoon sampler used in the standard penetration test, are kept in jars, tins or small polythene bags. Water samples should be taken whenever water is encountered during drilling. Samples are stored in jars whose lids are sealed by dipping them in paraffin wax. All samples must be clearly labelled, with labels both inside and outside the containers, and must be carefully transported and stored. Once they are no longer required for inspection or testing, samples may be discarded. However, care should be taken that they are not discarded too soon and all the people who may wish to make use of the samples should be informed before they are disposed of. In situ testing is carried out when:

i. Good quality sampling is impossible (for example, in granular soils, in fractured rock masses, in very soft or sensitive clays, or in stoney soils);

ii. The parameter required cannot be obtained from laboratory tests (for example, in situ horizontal stress);

iii. When in situ tests are cheap and quick, relative to the process of sampling and laboratory testing (for example, the use of the spt in clay, to determine undrained shear strength); and most importantly,

iv. For profiling and classification of soils (for example, with the cone test, or with dynamic penetration tests).

The most commonly used test is the Standard Penetration Test (SPT), which is routinely used at 1.5 m intervals within boreholes in granular soils, stoney soils, and weak rock. Other common in situ tests include the field vane (used only in soft and very soft cohesive soils), the plate test (used in granular soils and fractured weak rocks), and permeability tests (used in most ground, to determine the coefficient of permeability). The primary decision will be whether to test in the laboratory or in situ. Table 1.3 gives the relative merits of these options.


1-14 March 2009

Table 1.3 Relative Merits of In Situ and Laboratory Testing

In situ testing Laboratory testing

Advantages Test results can be obtained during the course of the investigation, much earlier than laboratory test results Appropriate methods may be able to test large volumes of ground, ensuring that the effects of large particle sizes and discontinuities are fully represented Estimates of in situ horizontal stress can be obtained

Tests are carried out in a well-regulated environment Stress and strain levels are controlled, as are drainage boundaries and strain rates Effective strength testing is straightforward The effect of stress path and history can be examined Drained bulk modulus can be determined

Disadvantages Drainage boundaries are not controlled, so that it cannot definitely be known whether loading tests are fully undrained Stress path and or or strain levels are often poorly controlled Tests to determine effective stress strength parameters cannot be made, because of the expense and inconvenience of a long test period Pore pressures cannot be measured in the tested volume, so that effective stresses are unknown.

Testing cannot be used whenever samples of sufficient quality and size are obtainable, for example, in granular soils, fractured weak rock, stoney clays Test results are only available some time after the completion of fieldwork

The ground investigation planner requires a detailed and up-to-date knowledge of both laboratory and in situ testing, if the best choices are to be made. Table 1.4 gives a summary of the local current situation — but this will rapidly become out of date. Whatever is used depends upon the soil and rock encountered, upon the need (profiling, classification, parameter determination), and upon the sophistication of geotechnical design that is anticipated.


March 2009 1-15

Table 1.4 Common Uses of In Situ and Laboratory Tests

Purpose Suitable laboratory test Suitable in situ test

Profiling Moisture content Particle size distribution Plasticity (Atterberg limits) Undrained strength

Cone test Dynamic penetration test Geophysical down-hole logging

Classification Particle size distribution Plasticity (Atterberg limits)

Cone

Parameter determination:

Undrained strength, cu Peak effective strength, c’ φ’ Residual strength, c’ φ’ Compressibility Permeability Chemical characteristics

Undrained triaxial Effective strength triaxial Shear box Ring shear Oedometer Triaxial, with small strain measurement Triaxial consolidation Triaxial permeability pH Sulphate content

SPT Cone Vane Self-boring pressuremeter Plate test In situ permeability tests Geophysical resistivity

The following table (Table 1.5 refers) details the applicable standards available for in-situ testing, while Table 1.6 details on standards available for laboratory soils testing.

Table 1.5 Standards Available for In Situ Testing

Test British Standard American StandardDensity tests (sand replacement, water replacement, core cutter,balloon and nuclear methods)

BS 1377: part 9: 1990, clause 2

ASTM D1556-82 ASTM D2937-83 ASTM D2937-84 ASTM D2922-91

Apparent resistivity BS 1377: part 9: 1990, clause 5.1 ASTM G57-78 (reapproved 1984)

In situ redox potential BS 1377: part 9: 1990, clause 5.2 In situ California bearing ratio BS 1377: part 9: 1990, clause 4.3 ASTM D4429-84 Standard penetration test BS 1377: part 9: 1990, clause 3.3 ASTM D1586-84

ASTM D4633-86 (energy measurement)

Dynamic penetration test BS 1377: part 9: 1990, clause 3.2 Cone penetration test BS 1377: part 9: 1990, clause 3.1 ASTM D3441-86 Vane test BS 1377: part 9: 1990, clause 4.4 ASTM D2573-72

(reapproved 1978) Plate loading tests BS 1377: part 9: 1990, clause

4.1, 4.2 ASTM D1194-72 (reapproved 1978) ASTM D4395-84

Pressuremeter test ASTM D4719-87


1-16 March 2009

Table 1.6 Standards Available for Laboratory Testing of Soils

Test British Standard American Standard

Classification tests Moisture content Atterberg limits Density Specific gravity Particle size distribution Pinhole dispersion test

BS 1377:part 2:1990, clause 3 BS 1377:part 2:1990, clause 4, 5 BS 1377:part 2:1990, clause 7 BS 1377:part 2:1990, clause 8 BS 1377:part 2:1990, clause 9

ASTM D2216-91 ASTM D4643-87 ASTM D4318-84 ASTM D854-92 ASTM D422-63 (reapproved 1972) ASTM D2217-85 ASTM D4647-87

Chemical tests Organic matter content Loss on ignition Sulphate content Carbonate content Chloride content pH Resistivity Redox potential

BS 1377:part 3:1990, clause 3 BS 1377:part 3:1990, clause 4 BS 1377:part 3:1990, clause 5 BS 1377:part 3:1990, clause 6 BS 1377:part 3:1990, clause 7 BS 1377:part 3:1990, clause 9 BS 1377:part 3:1990, clause 10 BS 1377:part 3:1990, clause 11

ASTM D2974-87 ASTM D4373-84 ASTM G51-77(reapproved 1984)

Compaction tests Proctor or 2.5kg rammer Heavy or 4.5kg rammer Vibrating hammer

BS 1377:part 4:1990, clause 3.3 BS 1377:part 4:1990, clause 3.5 BS 1377:part 4:1990, clause 3.7

ASTM D698-91 ASTM D1557-91

Strength tests California bearing ratio Undrained triaxial shear strength Effective strength from the consolidated-undrained triaxial compression test with pore pressure measurement Effective strength from the consolidated-drained triaxial compression test with volume change measurement Residual strength by direct shear testing in the shear box Residual strength using the Bromhead ring shear apparatus

BS 1377:part 4:1990, clause 7 BS 1377:part 7:1990, clause 8, 9 BS 1377:part 8:1990, clause 7 BS 1377:part 8:1990, clause 8 BS 1377:part 7:1990, clause 5

ASTM D1883-92 ASTM D2850-87 ASTM D3080-90

Compressibility tests One-dimensional compressibility in the oedometer Isotropic consolidation in the triaxial apparatus

BS 1377:part 5:1990, clause 3, 4 BS 1377:part 8:1990, clause 6

ASTM D2435-90

Permeability tests In the constant-head apparatus BS 1377:part 5:1990, clause 5 ASTM D2434-68

(reapproved 1974)


March 2009 1-17

The key points in checking the effectiveness of a site investigation are as follows. 1. Avoid excessive disturbance. Look for damaged cutting shoes, rusty, rough or dirty sample

barrels, or badly designed samplers. Check the depth of casings to ensure that these never penetrate beneath the bottom of the borehole. Try to assess the amount of displacement occurring beneath power augers, and prevent their use if necessary.

2. Check for water. Ensure that adequate water levels are maintained when drilling in granular soils or soft alluvium beneath the water table. The addition of water in small quantities should be kept to a minimum, since this allows swelling without going any way towards replacing total stress levels. Make sure the driller stops drilling when groundwater is met.

3. Check depths. The depths of samples can be found approximately by noting the number of rods

placed on the sampling tool as it is lowered down the hole, and the amount of ‘stick-up’ of the last rod at the top of the hole. This type of approach is often used by drillers, but is not always satisfactory. Immediately before any sample is taken or in situ test performed the depth of the bottom of the hole should be measured, using a weighted tape. If this depth is different from the last depth of the drilling tools then either the sides of the hole are collapsing, or soil is piping or heaving into the base. Open-drive sampling should not then be used.

4. Look for faulty equipment. On-site maintenance may lead to SPT hammers becoming non-

standard, for example owing to threading snapping and the central stem being shortened, giving a short drop. When working overseas with subcontract rigs the weight of the SPT hammer should also be measured. Other problems which often occur are: (i) the blocking of vents in sampler heads; and (ii) the jamming of inner barrels in double tube swivel-type corebarrels.

5. Examine driller’s records regularly. The driller should be aware that the engineer is seeking high

quality workmanship. One of the easiest ways of improving site investigation is to demand that up to the moment records are kept by the driller as drilling proceeds. These should then be checked several times a day when the engineer visits the borehole. Any problems encountered by the driller can then be discussed, and decisions taken.

1.7 METHODS OF SITE INVESTIGATION – DRILLING AND SAMPLING The next phase of the SI planning involves an appreciable understanding of the different methods commonly available for the local SI practices, and their corresponding use and limitations. This chapter briefly describes the equipment and procedures commonly used for the drilling and sampling of soil and rock. The methods addressed in this chapter are used to retrieve soil samples and rock cores for visual examination and laboratory testing. 1.7.1 Subsurface Exploration The primary functions of any ground investigation process will be one of the following:

i. Locating specific ‘targets’, such as dissolution features or abandoned mineworkings ii. Determining the lateral variability of the ground; iii. Profiling, including the determination of groundwater conditions; iv. Index testing; v. Classification; vi. Parameter determination.

1-18

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1-20 March 2009

• Barrel auger; and • Spiral auger. Figure 1.7 shows a selection of these augers.

Figure 1.7 Selection of Hand-Operated Augers 1.7.2.3 Wash Boring Wash boring is a relatively old method of boring small-diameter exploratory holes in fine-grained cohesive and non-cohesive soils. It was widely used in the USA in the first half of this century, but has been largely replaced by power auger methods. It is still used in areas of the world where labour is relatively cheap, for example southern Brazil. A very light tripod is erected, and a sheave is hung from it (Fig. 1.8). In its simplest form there are no motorized winches and the drilling water is pumped either by hand, or by a small petrol-driven water pump. Hollow drilling rods are connected to the pump via a flexible hose, and the drilling crew lift the string of rods by hand, or using a ‘cathead’ (a continuously rotating steel drum, around which a manilla rope is wound).


March 2009 1-21

Figure 1.8 Washboring Rig (Based On Hvorslev 1949) 1.7.3 Drilling Rotary drilling uses a rotary action combined with downward force to grind away the material in which a hole is being made. Rotary methods may be applied to soil or rock, but are generally easier to use in strong intact rock than in the weak weathered rocks and soils that are typically encountered during ground investigations. For a detailed description of equipment and methods the reader is referred to Heinz (1989). 1.7.3.1 Open-Holing Rotary methods may be used to produce a hole in rock, or they may be used to obtain samples of the rock while the hole is being advanced. The formation of a hole in the subsoil without taking intact samples is known as ‘open-holing’. It can be carried out in a number of ways, but in site investigation a commonly used tool is the ‘tricone rock roller bit’ (or roller core bit) (Fig. 1.9). 1.7.3.2 Coring The most common use of rotary coring in ground investigations is to obtain intact samples of the rock being drilled, at the same time as advancing the borehole. To do this a corebarrel, fitted with a ‘corebit’ at its lower end, is rotated and grinds away an annulus of rock. The stick of rock, the ‘core’, in the centre of the annulus passes up into the corebarrel, and is subsequently removed from the borehole when the corebarrel is full. The length of core drilled before it becomes necessary to remove and empty the corebarrel is termed a ‘run’.


1-22 March 2009

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1-24 March 2009

1.7.4 Exploration Pit Excavation Exploration pits and trenches permit detailed examination of the soil and rock conditions at shallow depths and relatively low cost. Exploration pits can be an important part of geotechnical explorations where significant variations in soil conditions occur (vertically and horizontally), large soil and or or non-soil materials exist (boulders, cobbles, debris) that cannot be sampled with conventional methods, or buried features must be identified and or or measured. Exploration pits are generally excavated with mechanical equipment (backhoe, bulldozer) rather than by hand excavation. The depth of the exploration pit is determined by the exploration requirements, but is typically about 2 m (6.5 ft) to 3 m (10 ft). In areas with high groundwater level, the depth of the pit may be limited by the water table. Exploration pit excavations are generally unsafe and or or uneconomical at depths greater than about 5 m (16 ft) depending on the soil conditions. 1.7.5 Probing A wide range of dynamic and static penetrometers are available, with different types being used in different countries. However, the objective of all probing is the same, namely to provide a profile of penetration resistance with depth, in order to give an assessment of the variability of a site. Probing is carried out rapidly, with simple equipment. It produces simple results, in terms of blows per unit depth of penetration, which are generally plotted as blowcount or depth graphs 1.7.5.1 MacKintosh Probe The Mackintosh prospecting tool (also commonly known as JKR probe) consists of rods which can be threaded together with barrel connectors and which are normally fitted with a driving point at their base, and a light hand-operated driving hammer at their top (Fig. 1.10). The tool provides a very economical method of determining the thickness of soft deposits such as peat. The driving point is streamlined in longitudinal section with a maximum diameter of 27mm. The drive hammer has a total weight of about 5kg. The rods are 1.2 m long and 12mm dia. The device is often used to provide a depth profile by driving the point and rods into the ground with equal blows of the full drop height available from the hammer: the number of blows for each 300 mm of penetration is recorded. When small pockets of stiff clay are to be penetrated, an auger or a core tube can be substituted for the driving point. The rods can be rotated clockwise at ground level by using a box spanner and tommy bar. Tools can be pushed into or pulled out of the soil using a lifting or driving tool. Because of the light hammer weight the Mackintosh probe is limited in the depths and materials it can penetrate. In Malaysia, this method of investigation is usually employed during preliminary investigative works. It involves the use of: • 5 kg hammer weight, • Dropped from a guided free fall height of 280mm (JKR probe), and • Usually carried out up to a depth of 12m, or upon encountering the 400 resistance blows or 300

mm.

The test itself is relatively cheap and quick to execute, and is used to establish: • Localised soft area or weak layer or spot or slip plane; • The presence of hard or bearing layers or shallow bedrocks, as in the case of limestone profiling; • Preliminary subsoil information (eg. soil consistency & undrained shear strength, cu); and • The interpolation between boreholes or piezocones.


March 2009 1-25

Limitations associated with this test include: • Relatively shallow test depths (deeper depths in coarse materials give misleading results); and • Prone to human errors: variation in drop weight or exerting force, gives rise to misleading

results, and risks of wrong counting unless mechanical counter is used. Precautionary measures to be observed require that: • The drop of the hammer should be free falling and consistent with each drop height; and • The components and apparatus must be properly washed and oiled.

Figure 1.11 Mackintosh Probe 1.7.6 Examination In-Situ 1.7.6.1 Trial Pit Trial pits provide the best method of obtaining very detailed information on strength, stratification, pre-existing shear surfaces, and discontinuities in soil. Very high quality block samples can be taken only from trial pits. It is as well to note that every year many people are killed during the collapse of unsupported trenches. Remember to be careful — do not enter trenches or pits more than 1.2m deep without either supporting the sides or battering back the sides. Even so, if a pit is dug and remains stable without support, a quick means of exit such as a ladder should be provided. Trial pits may be excavated by either hand digging or machine excavation. In general, machine excavation is used for shallow pits, whereas hand excavation is used for deep pits which must be supported. In the limited space of a trial pit, which is often 1.5m x 3m in plan area at ground level, it is usually impossible to place supports as machine excavation proceeds. Shallow trial pits provide a


1-26 March 2009

cheap method of examining near-surface deposits in situ, but the cost increases dramatically with depth, because of the need to support. Shallow trial pits can be excavated by wheeled offset backhoe which has a digging depth of about 3.5 – 4.0m, and may not be able to move easily across wet steeply sloping sites. Deeper pits, or pits where access is difficult can be excavated by 360° slew-tracked hydraulic excavators. These machines have a digging depth of about 6 m, and an available digging force about 50—100% greater than the backhoe.


March 2009 1-27

REFERENCES [1] Acker, W. L., III (1974). Basic Procedures for Soil Sampling and Core Drilling, Acker Drill Co. Inc., P.O. Box 830, Scranton, PA., 18501. [2] ADSC (1995). “Recommended procedures for the entry of drilled shaft foundations excavations.” The International Association of Foundation Drilling, (IAFD-ADSC), Dallas. [3] Contract DACW39-86-M-4273, Department of the Army, U.S. Army Corps of Engineers, Washington, D.C. [4] Hunt, R. E. (1984). Geotechnical Engineering Investigation Manual, McGraw-Hill Inc., 983 p. [5] Leroueil, S. and Jamiolkowski, M. (1991). “Exploration of soft soil and determination of design parameters”, Proceedings, GeoCoast’91, Vol. 2, Port & Harbor Res. Inst., Yokohama, 969-998. [6] Lowe III, J., and Zaccheo, P.F. (1991). "Subsurface explorations and sampling." Foundation Engineering Handbook, H. Y. Fang, ed., Van Nostrand Reinhold, New York, 1-71. [7] Lutenegger, A. J., DeGroot, D. J., Mirza, C., and Bozozuk, M. (1995). “Recommended guidelines for sealing geotechnical exploratory holes.” FHWA Report 378, Federal Highway Administration Washington, D.C. [8] Skempton, A. W. (1957). Discussion on “The planning and design of new Hong Kong airport.” Proceedings, Institution of Civil Engineers, Vol. 7 (3), London, 305-307. [9] U.S. Department of the Interior, Bureau of Reclamation. (1973). Design of small dams, United States Government Printing Office, Washington, D.C. [10] U.S. Department of the Interior, Bureau of Reclamation (1960). Earth manual, United States Government Printing Office, Washington, D.C.


1-28 March 2009


CHAPTER 2 SAMPLING AND SAMPLING DISTURBANCE


March 2009 2-i

Table of Contents


List of Table ........................................................................................................................... 2-ii


2.1 INTRODUCTION .......................................................................................................... 2-1

2.2 SAMPLING METHODS ................................................................................................... 2-1

2.2.1 Undistured Sample ........................................................................................ 2-1

2.2.2 Disturbed Sampling ....................................................................................... 2-4

2.3 SAMPLING INTERVAL AND APPROPRIATE SAMPLER TYPE ............................................... 2-5

2.4 SAMPLE RECOVERY ..................................................................................................... 2-5

2.5 REQUIRED VOLUME OF MATERIAL FOR TESTING PROGRAMME ...................................... 2-5

2.6 SAMPLE DISTURBANCE ................................................................................................ 2-7

REFERENCES ....................................................................................................................... 2-10


2-ii March 2009

List of Table


2.1 Common Sampling Methods 2-2

2.2 Mass of Disturbed Soil Sample Required For Various Tests 2-7


2.1 Effects of Tube Sampling Disturbance of Lightly Overconsolidated Natural (‘Structured’) 2-8

2.2 Influence of Tube Sampling Disturbance on Undrained Strength and Stiffness 2-9


March 2009 2-1

2 SAMPLING AND SAMPLING DISTURBANCE

2.1 INTRODUCTION Sampling is soil and rock is carried out for identification and description of soils strata and rock type with depth, and to perform laboratory testing for determination of index, classification and engineering properties. Laboratory tests typically consist of:

i. Index tests (for example, unconfined compressive strength tests on rock); ii. Classification tests (for example, Atterberg limit tests on clays); and iii. Tests to determine engineering design parameters (for example strength, compressibility,

and permeability). Samples obtained either for description or testing should be representative of the ground from which they are taken. They should be large enough to contain representative particle sizes, fabric, and fissuring and fracturing. They should be taken in such a way that they have not lost fractions of the in situ soil (for example, coarse or fine particles) and, where strength and compressibility tests are planned, they should be subject to as little disturbance as possible. 2.2 SAMPLING METHODS Generally, sampling during a soil investigation program can be grouped into two main categories.

1. Undisturbed sampling 2. Disturbed sampling

The methods of sampling adopted for a particular site investigation program is based on the type and requirement of soil investigation data for design and construction. While a large number of samplers and sampling methods are available, however, before a suitable technique can be selected it is always necessary to consider whether the sample size will be adequate, and whether the most suitable method of sampling has been selected, to ensure that sample disturbance is sufficiently small. 2.2.1 Undistured Sample Undisturbed samples are obtained with specialized equipment designed to minimize the disturbance to the in-situ structure and moisture content of the soils. The term “undisturbed” soil sample refers to the relative degree of disturbance to the soil’s in-situ properties. Specimens obtained by undisturbed sampling methods are used to determine the strength, stratification, permeability, density, consolidation, dynamic properties, and other engineering characteristics of soils. Undisturbed samples are obtained in clay soil strata for use in laboratory testing to determine the engineering properties of those soils. Undisturbed samples of granular soils can be obtained, but often specialized procedures are required such as freezing or resin impregnation and block or core type sampling. Common methods for obtaining undisturbed samples are summarized in Table 2.1 and briefly described below.


2-2 March 2009

Table 2.1 Common Sampling Methods

Sampler Disturbed/

Undisturbed Appropriate Soil Types Method of Penetration % Use in

Practice Split-Barrel

(Split Spoon) Disturbed Sands, silts, clays Hammer driven 85

Thin-Walled Shelby Tube

Undisturbed Clays, silts, fine-grained soils, clayey sands

Mechanically Pushed 6

Continuous Push

Partially Undisturbed

Sands, silts, and clays Hydraulic push with plastic lining

4

Piston Undisturbed Silts and clays Hydraulic push 1 Pitcher Undisturbed Stiff to hard clay, silt,

sand, partially weathered rock and frozen or resin

impregnated granular soil

Rotation and hydraulic pressure

<1

Denison Undisturbed Stiff to hard clay, silt, sand and partially

weather rock

Rotation and hydraulic pressure

<1

Modified California

Disturbed Sands, silts, clays and gravels

Hammer driven (large split spoon)

<1

Continuous Auger

Disturbed Cohesive soils Drilling w/ Hollow Stem Augers

<1

Bulk Disturbed Gravels, sands, silts, clays Hand tools, bucket augering

<1

Block Undisturbed Cohesive soils and frozen or resin impregnated

granular soils

Hand tools <1

Brief description of some of the undisturbed samplers and sampling method is presented below: Thin Walled Shelby Tube It is used in cohesive soil in rotary drilled boreholes. The sampler consists of a thin walled tube secured to a head containing a ball check valve. The most commonly used sampler sizes are 75mm and 50 mm (OD). The head is threaded to a drill rod. The tube is used below the casing after cleaning the casing side with clean out auger and washing the bottom. The sampler screwed to the drill rod and the sampler are pressed with hydraulic jack of the drilling in one fast stroke. The pressure required to push the sampler should be read on hydraulic gauge. It is always preferable to wait for some time say 15 to 30 minutes before pulling out the tube as this allows the soil to expand and gain shear strength so that it sticks to the tube. The soil is removed from the two ends of the tube to about 25mm to 50mm and sealed with paraffin wax. The scrapped sample is logged in the Borelog. Piston Sampler This is similar to the Shelby tube sampler except that it is sealed at the bottom and therefore contamination is avoided. Secondly by holding the piston stationary and pressing the tube downward, the top of the sample is protected against distortion. The most commonly used sampler size is 73.6mm dia and 762mm long sampler. The actuating rod is held in place while the sampling tube is pushed past the stationary piston. The tube is removed from the hole and separated from sampler by unscrewing the actuating rod. The tube ends are sealed. The sample should be logged in the borelog as done for Shelby tube.


March 2009 2-3

Denison Sampler The Denison sampler was designed by H.L. Johnson in 1930 to obtain samples from dense or highly cemented strata (stiff to hard clays and dense sand) where thin Shelby tube was unable to penetrate and extract an undisturbed sample. The sample device is essentially a double tube core barrel with thin lined liner tube adapted to soil use. The inner tube with cutting shoe always advances ahead of the rotating outer core barrel ensuring the sample to e undisturbed and uncontaminated. The Denison core barrel is manufactured in 89, 100, 140, and 197mm OD sizes, and recovers relatively large samples in the inner stationary tube. The standard lengths are 60cm and 1.5m. Besides stiff and dense soils, the sampler can also sample clean sand and soft clays with use of drilling mud, vacuum valve and basket core retainer. The operating procedure is to lower the sampler to the bottom of the hole and apply hydraulic feed downward pressure, simultaneously drilling at a maximum rate of 100 rpm and allowing the circulation of drilling fluid just enough to wash the cutting. Once the depth is reached, the core barrel is withdrawn, the head and cutting shoe is removed and inner liner pushed hydraulically or mechanically from inner core barrel. The soil sample collected in the liner is sealed in th same was as Shelby tube and the sample is logged. Pitcher Sampler The pitcher sampler is basically a Denison sampler in which the inner barrel is spring loaded so as to provide for the automatic adjustment of the distance by which the cutting edge of the barrel leads the coring bit. After cleaning the drill hole, the sampler is lowered to the bottom of the drill hole. When the sampler reaches the bottom of the drill hole the inner tube meets the resistance first and the outer barrel slides past the tube until the spring at the top of the tube contacts the top of the outer barrel. The spring in the sampler is compressed with respect to the amount of resistance met by the soil sample i.e soft or hard. Sampling is accomplished by rotating the outer barrel at 100 to 200 rpm while exerting the downward pressure. Upon completion of the sampling drive, the sampler is removed from the borehole, and the inner tube which is used to ship and store the sample is removed from the sampler. Mazier’s Sampler The Mazier’s sampler, commonly used in south-east Asia, for soil exploration is very much similar to Denison sampler. It is very useful for obtaining samples of stiff to hard residual soil with relict rock fragments and weathered material. The Mazier’s triple tube retractor barrel which is a stationary plastic liner encasing 73 m diameter core is compatible with standard laboratory and testing apparatus. The Mazier’s sample is used in conjunction with double core barrel when coring of rock is required. Block Samples For projects where the determination of the undisturbed properties is very critical, and where the soil layers of interest are accessible, undisturbed block samples can be of great value. Of all the undisturbed testing methods discussed in this manual, properly-obtained block samples produce samples with the least amount of disturbance. Such samples can be obtained from the hillsides, cuts, test pits, tunnel walls and other exposed sidewalls. Undisturbed block sampling is limited to cohesive soils and rocks. The procedures used for obtaining undisturbed samples vary from cutting large blocks of soil using a combination of shovels, hand tools and wire saws, to using small knives and spatulas to obtain small blocks. In addition, special down-hole block sampling methods have been developed to better obtain samples in their in-situ condition. For cohesive soils, the Sherbrooke sampler has been developed and is able to obtain samples 250 mm (9.85 in) diameter and 350 mm (13.78 in) height (Lefebvre


2-4 March 2009

and Poulin 1979). In-situ freezing methods for saturated granular soils and resin impregnation methods have been implemented to “lock” the soil in the in-situ condition prior to sampling. When implemented, these methods have been shown to produce high quality undisturbed samples. However, the methods are rather involved and time consuming and therefore have not seen widespread use in practice. Once samples are obtained and transported to the laboratory in suitable containers, they are trimmed to appropriate size and shape for testing. Block samples should be wrapped with a household plastic membrane and heavy duty foil and stored in block form and only trimmed shortly before testing. Every sample must be identified with the following information: project number, boring or exploration pit number, sample number, sample depth, and orientation. 2.2.2 Disturbed Sampling Disturbed samples are those obtained using equipment that destroy the macrostructure of the soil but do not alter its mineralogical composition. Specimens from these samples can be used for determining the general lithology of soil deposits, for identification of soil components and general classification purposes, for determining grain size, Atterberg limits, and compaction characteristics of soils. Disturbed samples can be obtained with a number of different methods as summarized in Table 2.1. Some of the sampling methods given in Table 2.1 are described below. Split-Barrel (Split Spoon) The split spoon sampler is a solid steel tube barrel split into two halves longitudinally. The device has a check valve and a hard steel shoe. When the head and shoe are unscrewed the barrel opens in the centre exposing the sample. Improvement in design provides liner and the sampler retainer. The ball valve in the head and the sample retainer valve spring prevent the sample particularly cohesionless soil from being washed out and lost. The borehole is cleaned before lowering the sampler into the borehole. The sampler is then driven into the borehole base by hammering to extract the sample. The sample is then logged on a borelog. Continuous Auger Continuous auger or continuous flight augers are augers with continuous spiral on the shaft. As the hole advances, additional sections of spiral flight are added. In this type of auger, the soils rise to the top of the hole on the spiral flight and is sampled as it emerges. Moreover the disadvantage of raising and lowering the auger to remove the soil is eliminated. Condinuous augers can be with solid or hollow stems also. The limitation of the augers is that these are not effective below water table as there are constant caving problems and samples are washed off unless cased. Hollow stem auger can cope with the situation to some extent with special adaptors. The limitations are maximum depth 30m for continuous augers. Bulk Samples Bulk samples are suitable for soil classification, index testing, R-value, compaction, California Bearing Ratio (CBR), and tests used to quantify the properties of compacted geomaterials. The bulk samples may be obtained using hand tools without any precautions to minimize sample disturbance. The sample may be taken from the base or walls of a test pit or a trench, from drill cuttings, from a hole dug with a shovel and other hand tools, by backhoe, or from a stockpile. The sample should be put into a container that will retain all of the particle sizes. For large samples, plastic or metal buckets or metal barrels are used; for smaller samples, heavy plastic bags that can be sealed to maintain the water content of the samples are used.


March 2009 2-5

Usually, the bulk sample provides representative materials that will serve as borrow for controlled fill in construction. Laboratory testing for soil properties will then rely on compacted specimens. If the material is relatively homogeneous, then bulk samples may be taken equally well by hand or by machine. However, in stratified materials, hand excavation may be required. In the sampling of such materials it is necessary to consider the manner in which the material will be excavated for construction. If it is likely that the material will be removed layer by layer through the use of scrapers, samples of each individual material will be required and hand excavation from base or wall of the pit may be a necessity to prevent unwanted mixing of the soils. If, on the other hand, the material is to be excavated from a vertical face, then the sampling must be done in a manner that will produce a mixture having the same relative amounts of each layer as will be obtained during the borrow area excavation. This can usually be accomplished by hand-excavating a shallow trench down the walls of the test pit within the depth range of the materials to be mixed. 2.3 SAMPLING INTERVAL AND APPROPRIATE SAMPLER TYPE In general, SPT samples are taken in both granular and cohesive soils, and thin-walled tube samples are taken in cohesive soils. The sampling interval will vary between individual projects and between regions. A common practice is to obtain split barrel samples at 0.75 m (2.5 ft) intervals in the upper 3 m (10 ft) and at 1.5 m (5 ft) intervals below 3 m (10 ft). In some instances, a greater sample interval, often 3 m (10 ft), is allowed below depths of 30 m (100 ft). In other cases, continuous samples may be required for some portion of the boring. In cohesive soils, at least one undisturbed soil sample should be obtained from each different stratum encountered. If a uniform cohesive soil deposit extends for a considerable depth, additional undisturbed samples are commonly obtained at 3 m (10 ft) to 6 m (10 ft) intervals. Where borings are widely spaced, it may be appropriate to obtain undisturbed samples in each boring; however, for closely spaced borings, or in deposits which are generally uniform in lateral extent, undisturbed samples are commonly obtained only in selected borings. In erratic geologic formations or thin clay layers it is sometimes necessary to drill a separate boring adjacent to a previously completed boring to obtain an undisturbed sample from a specific depth which may have been missed in the first boring. 2.4 SAMPLE RECOVERY Occasionally, sampling is attempted and little or no material is recovered. In cases where a split barrel or another disturbed-type sample is to be obtained, it is appropriate to make a second attempt to recover the soil sample immediately following the first failed attempt. In such instances, the sampling device is often modified to include a retainer basket, a hinged trap valve, or other measures to help retain the material within the sampler. In cases where an undisturbed sample is desired, the field supervisor should direct the driller to drill to the bottom of the attempted sampling interval and repeat the sampling attempt. The method of sampling should be reviewed, and the sampling equipment should be checked to understand why no sample was recovered (such as a plugged ball valve). It may be appropriate to change the sampling method and/or the sampling equipment, such as waiting a longer period of time before extracting the sampler, extracting the sampler more slowly and with greater care, etc. This process should be repeated or a second boring may be advanced to obtain a sample at the same depth. 2.5 REQUIRED VOLUME OF MATERIAL FOR TESTING PROGRAMME A further consideration in fixing sample sizes is the standard test specimen sizes in use. The specimen sizes commonly used here and in United Kingdom is shown below.


2-6 March 2009

Compressibility characteristics Oedometer 76mm dia. x 19mm high Triaxial cell 102mm dia. x 102mm high Hydraulic consolidation cell up to 254mm dia. x 100 – 125mm high Triaxial compression tests Small specimens 38mm dia. x 76mm high Large specimens 102mm dia. x 204mm high or 152mm dia. x 305mm high Direct shear tests Small specimens 60mm x 60mm in plan Large specimens 305mm x 305mm in plan

Small triaxial specimens are normally tested in groups of three, all of which should be obtained from the same level in the sample in order that they are as similar as possible. Three 38mm dia. Specimens can be obtained from a 102 mm dia. sample. Soil testing equipment manufactured in the USA uses the following specimen sizes.

Compressibility characteristics Consolidometer (large specimen) 113mm dia. (small specimen) 64mm dia. Triaxial compression tests Small specimens 36mm dia. x 71mm high Medium specimens 71mm dia x 142mm high Large specimens 102mm dia. x 204mm high Or 152mm dia. x 305mm high Direct shear tests Cylindrical specimens 63.5mm dia. Square specimens 51mm dia. x 52mm

Three 36mm dia. (1.4in. dia.) specimens can be obtained from either 89mm (3.5 in.) dia. samples or 102 mm (4 in.) dia. samples. As noted above, when discussing the need for samples to contain representative particle sizes, in many cases it is the minimum quantity of soil required for a particular test procedure which will dictate the volume or mass that must be obtained. BS 5930: 1981 suggested sample sizes should be determined on the basis both of soil type and the purpose for which the sample was needed (Table 2.2).


March 2009 2-7

Table 2.2 Mass of Disturbed Soil Sample Required For Various Tests

Testing

Soil type

Clay, silt or sand (kg)

Fine and medium gravel (kg)

Coarse gravel (kg)

Moisture content, Atterberg limits, sieve analysis, chemical tests

1 5 30

Compaction tests 25-60 25-60 25-60 Soil stabilization tests 100 130 160

(Source: BS 5930: 1981) 2.6 SAMPLE DISTURBANCE The most obvious effect of sample disturbance can be seen when attempting to tube sample very soft, sensitive clays with a poorly designed sampler. The soil around the edge of the sample undergoes a very large decrease in strength, such that when the tube is withdrawn from the soil there is no recovery. But, as has been noted above, sample disturbance occurs in all sampling processes and, if sampling is carried out well, the effects of disturbance will hopefully be more subtle. Whatever its magnitude, sampling disturbance normally affects both undrained strength and compressibility. In addition, chemical effects may cause changes in the plasticity and sensitivity of the soil sample. (I) Strength Although it has been noted above that tube sampling disturbance has the greatest effect, in terms of reductions in mean effective stress, on reconstituted clays its effect on the undrained shear strength of such material is, perhaps surprisingly, small. Laboratory experiments by a number of workers have shown that the stress paths during undrained shearing converge on the critical state and, because the soil is initially reconstituted, the state boundary surface is not disrupted by tube sampling. Typically, it has been found that the undrained strength is reduced by less than 10%, even when the material is not reconsolidated back to its initial stress state (for example, Siddique (1990)). Tube sampling does, however, have a significant effect on real soils, most of which are either bonded (‘structured’), and/or more heavily overconsolidated. Shearing of bonded soils during tube sampling can have the effect of progressively destructuring them. Clayton et al. (1992) show comparisons of the stress paths taken by soil specimens tube sampled in different ways. Figure 2.1 shows how tube sampling a lightly overconsolidated natural, structured clay with a standard piston sampler leads subsequently to much higher pore pressure generation during undrained shear, with the consequence that undrained strength is reduced. Clayton et al. (1992) found that provided tube sampling strain excursions were limited to ± 2% and that appropriate stress paths were used to reconsolidate the material back to its in situ stress state, the undrained strength of the Bothkennar clay would be within ± 10% of its undisturbed value. It is to be expected, however, that much greater effects will occur when sensitive clays are sampled. Heavily overconsolidated clays often display almost vertical stress paths under undrained shear. An increase in the mean effective stress level as a result of tube sampling will result in approximately proportional increase in intact strength. Unfortunately, however, this is not the only effect at work. Hammering of tubes into stiff clays can cause fracturing, and loosening along fissures, and this may lead to a marked reduction in measured undrained strength.

2-8

Figure

(II) C The effecto assessaxial strareliable stests, It geotechnmean effresult of Because measurewill avoidwith respthe stiffn

2.1 Effects ofClay on: (

(b)

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March 20

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March 200

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2-10 March 2009

REFERENCES [1] Acker, W. L., III (1974). Basic Procedures for Soil Sampling and Core Drilling, Acker Drill Co. Inc., P.O. Box 830, Scranton, PA., 18501. [2] American Association of State Highway and Transportation Officials (AASHTO) (1988). Manual on Subsurface Investigations, Developed by the Subcommittee on Materials, Washington, D.C. [3] American Association of State Highway and Transportation Officials (AASHTO). (1995). Standard specifications for transportation materials and methods of sampling and testing: part II: tests, Sixteenth Edition, Washington, D.C. [4] Deere, D. U. (1963). “Technical description of rock cores for engineering purposes.” Felmechanik und Ingenieur Geologis, 1 (1), 16-22. [5] Ford, P.J., Turina, P.J., and Seely, D.E. (1984). Characterization of hazardous waste sites - a methods manual: vol. II, available sampling methods, 2nd Edition, EPA 600/4-84-076 (NTIS PB85-521596). Environmental Monitoring Systems Laboratory, Las Vegas, NV. [6] Hunt, R. E. (1984). Geotechnical Engineering Investigation Manual, McGraw-Hill Inc., 983 p. Hvorslev, M. J. (1948). Subsurface Exploration and Sampling of Soils for Civil Engineering Purposes, U.S. Army Corps of Engineers, Waterways Experiment Station, Vicksburg, MS. [7] Krebs, R. D., and Walker, E. D. (1971). "Highway materials." Publication 272, Department of Civil Engrg., Massachusetts Institute of Technology, McGraw-Hill Company, New York, 107. [8] Kulhawy, F.H., Trautmann, C.H., and O’Rourke, T.D. (1991). “The soil-rock boundary: What is it and where is it?” Detection of and Construction at the Soil/Rock Interface, GSP No. 28, ASCE, Reston/VA, 1-15. [9] Kulhawy, F.H. and Phoon, K.K. (1993). “Drilled shaft side resistance in clay soil to rock”, Design and Performance of Deep Foundations: Piles & Piers in Soil & Soft Rock, GSP No. 38, ASCE, Reston/VA, 172-183. [10] Leroueil, S. and Jamiolkowski, M. (1991). “Exploration of soft soil and determination of design parameters”, Proceedings, GeoCoast’91, Vol. 2, Port & Harbor Res. Inst., Yokohama, 969-998. [11] Lowe III, J., and Zaccheo, P.F. (1991). "Subsurface explorations and sampling." Foundation Engineering Handbook, H. Y. Fang, ed., Van Nostrand Reinhold, New York, 1-71. [12] Lupini, J.F., Skinner, A.E., and Vaughan, P.R. (1981). "The drained residual strength of cohesive soils", Geotechnique, Vol. 31 (2), 181-213. [13] Lutenegger, A. J., DeGroot, D. J., Mirza, C., and Bozozuk, M. (1995). “Recommended guidelines for sealing geotechnical exploratory holes.” FHWA Report 378, Federal Highway Administration Washington, D.C. [14] NAVFAC, P-418. (1983). "Dewatering and groundwater control." Naval Facilities Engineering Command, Department of the Navy; Publication No. TM 5-818-5. [15] Powers, J. P. (1992). Construction Dewatering, John Wiley & Sons, Inc., New York.


March 2009 2-11

[16] U.S. Environmental Protection Agency (EPA). (1991). Description and sampling of contaminated soils, (EPA/625/12-9/002; November), Washington, D.C. [17] U.S. Department of the Interior, Bureau of Reclamation. (1973). Design of small dams, United States Government Printing Office, Washington, D.C. [18] U.S. Department of the Interior, Bureau of Reclamation (1960). Earth manual, United States Government Printing Office, Washington, D.C


2-12 March 2009


CHAPTER 3 IN SITU GEOTECHNICAL TESTING

Chapter 3 IN-SITU GEOTECHNICAL TESTING

March 2009 3-i

Table of Contents


List of Table ........................................................................................................................... 3-ii


3.1 INTRODUCTION .......................................................................................................... 3-1

3.1 STANDARD PENETRATION TEST (SPT) .......................................................................... 3-1

3.1.1 Correction Factors for Spt .............................................................................. 3-4

3.2 CONE PENETRATION TEST (CPT) .................................................................................. 3-5

3.3 FIELD VANE SHEAR TEST (VST) ................................................................................... 3-15

3.4 SUMMARY ON IN-SITU GEOTECHNICAL METHODS ........................................................ 3-20

3.5 GROUNDWATER INVESTIGATIONS .............................................................................. 3-21

3.5.1 General ....................................................................................................... 3-21

3.5.2 Determination of Ground Water Levels and Pressures ..................................... 3-22

3.5.3 Field Measurement of Permeability ................................................................ 3-22

REFERENCES ....................................................................................................................... 3-24


3-ii March 2009

List of Table


3.1 Comparison between Advantages and Disadvantages of SPT 3-4

3.2 Comparison between Advantages and Disadvantages in CPT 3-5

3.3 Diagnostic Features of Soil Type 3-14

3.4 General Advantages and Disadvantages of VST 3-16

3.5 Field Methods for Measurement of Permeability 3-23

List of Figures


3.1 Common In-Situ Tests for Geotechnical Site Characterization of Soils 3-1

3.2(a) Equipment for the Standard Penetration Test 3-2

3.3 Ratio of Undrained Shear Strength (Cu) Determined On 100mm Diameter. 3-4

3.4 Original Dutch Cone and Improved Mechanical Delft Cone (Lousberg Et Al. 1974) 3-6

3.5 Begemann Mechanical Friction Cone (Left, Fully Closed; Right, Fully Extended) 3-7

3.6 Electric Friction Cone (Largely After Meigh 1987) 3-8

3.7 Definition of Cone Area Ratio, Α 3-9

3.8 Distribution of Excess Pore Pressure over the Cone (Coutts 1986). 3-10

3.9 Typical Record of a Friction Cone Penetration Test (Te Kamp, 1977, from Meigh, 1987) 3-12

3.10 (a) relationship between soil type, cone resistance and local friction (Begemann 1956); 3-13

3.11 Ratio of (CPT Qc) (SPT N) as a Function of D50 Particle Size of the Soil (Thorburn, 1971). 3-14

3.12 General Test Procedures for the Field Vane in Fine-Grained Soils. 3-16

3.13 Assumed Geometry of Shear Surface for Conventional Interpretation of the Vane Test 3-18

3.14 Vane Correction Factor (:R) Expressed in Terms of Plasticity Index and Time to Failure. 3-20

3.15 Relevance of In-Situ Tests to Different Soil Types 3-21

3.16 Field Permeability Test Arrangement for Soil 3-23


March 2009 3-1

3 IN-SITU GEOTECHNICAL TESTING

3.1 INTRODUCTION Several in-situ tests define the geostratigraphy and obtain direct measurements of soil properties and geotechnical parameters. The common tests include: standard penetration test (SPT), cone penetration test (CPT), piezocone test (CPTu), flat dilatometer test (DMT), borehole pressure meter test (PMT), and vane shear test (VST). Each test applies different loading schemes to measure the corresponding soil response in an attempt to evaluate material characteristics, such as strength and/or stiffness. Fig. 3.1 depicts these various devices and simplified procedures in graphical form. Details on these tests will be given in the subsequent sections.

Figure 3.1 Common In-Situ Tests for Geotechnical Site Characterization of Soils

Boreholes are required for conducting the SPT and normal versions of the PMT and VST. A rotary drilling rig and crew are essential for these tests. In the case of the CPT, CPTU, and DMT, no boreholes are needed, thus termed direct-push technologies. Specialized versions of the PMT (i.e., full-displacement type) and VST can be conducted without boreholes. As such, these may be conducted using either standard drill rigs or mobile hydraulic systems (cone trucks) in order to directly push the probes to the required test depths. A disadvantage of direct-push methods is that hard cemented layers and bedrock will prevent further penetration. In such cases, borehole methods prevail as they may advance by coring or non-coring techniques. An advantage of direct-push soundings is that no cuttings or spoil are generated. 3.1 STANDARD PENETRATION TEST (SPT) The standard penetration test (SPT) is performed during the advancement of a soil boring to obtain an approximate measure of the dynamic soil resistance, as well as a disturbed drive sample (split barrel type). The test was introduced by the Raymond Pile Company in 1902 and remains today as the most common in-situ test worldwide.

3-2

The SPTnumber weight sym (30 iincremenincremenm or blow The pene300mm penetratthe hole.

T involves theof blows to aystem is usedinches) to ant is recordednts are summws per foot).

etration resis(1 ft) of peion is ignored.

Chap

e driving of advance the sd for the pounchieve threed as a .seatin

med to give thFigs. 3.2 a, b

tance (N) is tenetration. Td, because th

Figure 3.2(a)

pter 3 IN-SITU

a hollow thisplit-barrel sanding where

e successive ng, while thehe N-value ("bb refer.

the number oThe penetratie soil is cons

Equipment fo

GEOTECHNICA

ick-walled tuampler a vertia 63.5-kg (14increments

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of blows requion resistanceidered to hav

or the Standa

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on Test

March 20

measuring th1 foot). A drofalls from 0.7

each. The firecond and thied in blows/0

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009

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ast of ng


March 2009 3-3

Figure 3.2(b) Sequence of Driving Split-Barrel Sampler During the Standard Penetration Test Correlations between SPT N value and soil (Fig. 3.3 refers) or weak rock properties are wholly empirical, and depend upon an international database of information. Because the SPT is not completely standardised, these correlations cannot be considered particularly accurate in some cases, and it is therefore important that users of the SPT and the data it produces have a good appreciation of those factors controlling the test, which are: 1. Variations in the test apparatus; 2. The disturbance created by boring the hole; and 3. The soil into which it is driven.


3-4 March 2009

Figure 3.3 Ratio of Undrained Shear Strength (Cu) Determined On 100mm Diameter. Specimens to SPT N, As a Function of Plasticity (Stroud 1974).

A comparison between advantages and disadvantages of SPT is summarised in Table 3.1 as follows:

Table 3.1 Comparison between Advantages and Disadvantages of SPT

Advantages Disadvantages Simple and rugged Disturbed sample (index tests only) Suitable in many soil types Crude number for analysis Can perform in weak rocks Not applicable in soft clays and silts Easily available High variability and uncertainty

3.1.1 Correction Factors for Spt In recent years, it has become a practice to adjust the N valule of SPT test by a hammer-energy ratio or hammer efficiency of 60% and much attention has been given to N values because of its use in liquefaction studies. Geotechnical foundation practice and engineering usage based on SPT correlations have been developed on the basis of standard-of-practice corresponding to an average ER = 60 %. Normally the correction factor used for SPT tests N values is

(N1)60 = N.CN.CE (3.1)

Where (N1)60 = Corrected N Value N = SPT N count obtained from Testing

10

8

6

4

2

0 0 10 20 30 40 50 60 70

C u/N

(kN

/m2 )

PI %

Boulder clay Kimmeridge clayLaminated clay Woolwich and Reading clay Sunnybrook till Upper Lias clay London clay Keuper marl Bracklesham bods Flints Oxford clay


March 2009 3-5

CN. = Depth Correction Factor - (Should not be greater than 1.7)

= (1/σ’vo)0.5 - (Liao and Whitman 1986) = 2.2/ (1.2 + σ’vo / Pa) - (Kayen et. Al, 1992) σ’vo = Effective overburden pressure (γt - γw).z in tons / sq ft Pa = 1 tons/sq. ft (95KN/m2) CE = Correction Factor for Energy Ratio of 60%. = ER /60

ER = Energy Ratio for drill rigs (Table below) Additional correction has been proposed by (Skempton, 1986, Robertson and Wride, 1998) for hammer type (donut and safety), borehole diameter rod lengths and sampler. 3.2 CONE PENETRATION TEST (CPT) The cone penetration test is quickly becoming the most popular type of in-situ test because it is fast, economical, and provides continuous profiling of geostratigraphy and soil properties evaluation. The CPT can be used in very soft clays to dense sands, yet is not particularly appropriate for gravels or rocky terrain. The pros and cons are listed in Table 3.2 below. As the test provides more accurate and reliable numbers for analysis, yet no soil sampling, it provides an excellent complement to the more conventional soil test boring with SPT measurements.

Table 3.2 Comparison between Advantages and Disadvantages in CPT

Advantages Disadvantages Fast and continuos profiling High capital investment Economical and productive Requires skilled operator to run Results not operator-dependent Electronic drift, noise and calibration Strong theoretical basis in interpretation No soil samples are obtained Particularly suitable for soft soils Unsuitable for gravel or boulder deposits

except where special rigs are provided and / or additional drilling support is available.

Samples of various cone penetrometers are illustrated in Figs. 3.4, 3.5 and 3.6.

Country Hammer Releases ER (%) USA Safety 2 turns of Rope 55 Donut 2 turns of Rope 45 Japan Donut Tombi 78 -85 Donut 2 turns of Rope 65 – 67 China Automatic Trip 60 Donut Manual 55 UK Automatic Trip 73

3-6

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March 20

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March 200

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March 20

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March 200

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3-10

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Figure 3.8

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March 20

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March 2009 3-11

where qn = net cone resistance, and σv = vertical total stress at the level at which qn is measured. Net cone resistance can only be calculated once the distribution of bulk unit weight with depth is known, or can be estimated. Typical results of a friction cone test are given in Fig. 3.9. The original development of side friction measurement was made by Begemann using a mechanical cone, who found the useful correlation between friction ratio and soil type shown in Fig. 3.10a. He defined soil type by its percentage of particles finer than 0.016mm, and found that on a plot of side friction versus cone resistance each type of soil plotted as a straight line passing through the origin. This has led to more sophisticated charts such as that shown in Fig. 3.10b, and for the piezocone to correlations based upon the relationship between excess pore pressure and net cone resistance (qn = qc - σv).

3-12

Figure

3.9 Typical R

Chap

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GEOTECHNICA

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March 20

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March 200

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March 20

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March 2009 3-15

In a comparative study based upon case records, Dikran found that the ratio of calculated/observed settlements fell in the range 0.21—2.72, for four traditional methods of calculation using the CPT. For the SPT the variation was 0.15—10.8. When calculating the point resistance of piles in sand based upon cone resistance, it is normal to consider the static cone penetrometer as a model of the pile, and simply apply a reduction factor of between two and six to give allowable bearing pressure (Van der Veen and Boersma 1957; Sanglerat 1972). Sand deposits are rarely uniform, and so an averaging procedure is used with the qc values immediately above and below the proposed pile tip position (Schmertmann 1978). The side friction of piles may be calculated directly from the side friction of the cone, or by correlation with cone resistance. In cohesive soils, the CPT is routinely used to determine both undrained shear strength and compressibility. In a similar way to the bearing capacity of a foundation, cone resistance is a function of both overburden pressure (σv) and undrained shear strength (cu): qc = NkCu+σv (3.8) so that the undrained shear strength may be calculated from:

cu = qc-σv

Nk (3.9)

provided that Nk is known, or can be estimated. The theoretical bearing capacity factor for deep foundation failure cannot be applied in this equation because the cone shears the soil more rapidly than other tests, and the soil is failed very much more quickly than in a field situation such as an embankment failure. At shallow depths, or in heavily over-consolidated soils, the vertical total stress in the soil is small, so that: cu qc

Nk (3.10)

Typically, in these conditions, the undrained shear strength is about 1/15th to 1/20th of the cone resistance. Nk is not a constant, but depends upon cone type, soil type, overconsolidation ratio, degree of cementing, and the method by which undrained shear strength has been measured (because undrained shear strength is sample-size and test-method dependent). The Nk value in an over-consolidated clay will be higher than in the same clay when normally consolidated Typically, Nk varies from 10 to 20. Lunne and Kleven have shown that this variation is significantly reduced, giving Nk much closer on average to 15, if a correction (Nk* = Nk/µ) is made to allow for rate effects, in a similar way to that proposed by Bjerrum for the vane test (see below), but this is rarely done in practice. 3.3 FIELD VANE SHEAR TEST (VST) The vane shear test (VST), or field vane (FV), is used to evaluate the in-place undrained shear strength (suv) of soft to stiff clays & silts at regular depth intervals of 1 meter (3.28 feet). The test consists of inserting a four-bladed vane into the clay and rotating the device about a vertical axis. Limit equilibrium analysis is used to relate the measured peak torque to the calculated value of su. Both the peak and remoulded strengths can be measured; their ratio is termed the sensitivity, St. A


3-16 March 2009

selection of vanes is available in terms of size, shape, and configuration, depending upon the consistency and strength characteristics of the soil. The standard vane has a rectangular geometry with a blade diameter D = 65 mm, height H = 130 mm (H/D =2), and blade thickness e = 2 mm. Fig. 3.12 illustrates the general VST procedures

Figure 3.12 General Test Procedures for the Field Vane in Fine-Grained Soils. (Note: Interpretation of Undrained Strength Shown Is Specifically For Standard Rectangular Vane with H/D = 2)

The general advantages and disadvantages of VST is summarised in Table 3.4 as follows.

Table 3.4 General Advantages and Disadvantages of VST

Advantages Disadvantage Assessment of undrained strength, suv Limited application to soft to stiff clays Simple test and equipment Slow and time-consuming Measure in-situ clay sensitivity (St) Raw suv needs (empirical) correction Long history of use in practice Can be affected by sand lenses and seams

By implication, BS 1377 considers that the field vane will not be suitable for testing soils with undrained strengths greater than about 75 kPa. The vane must be designed to achieve an area ratio of 12% or less. The test is not suitable for fibrous peats, sands or gravels, or in clays containing laminations of silt or sand, or stones.


March 2009 3-17

Interpretation The vane test is routinely used only to obtain ‘undisturbed’ peak undrained shear strength, and remoulded undrained shear strength. The undrained strength is derived on the basis of the following assumptions: 1. Penetration of the vane causes negligible disturbance, both in terms of changes in effective

stress, and shear distortion; 2. No drainage occurs before or during shear; 3. The soil is isotropic and homogeneous; 4. The soil fails on a cylindrical shear surface; 5. The diameter of the shear surface is equal to the width of the vane blades; 6. At peak and remoulded strength there is a uniform shear stress distribution across the shear

surface; and 7. There is no progressive failure, so that at maximum torque the shear stress at all points on

the shear surface is equal to the undrained shear strength, cu. On this basis (Fig. 3.13), the maximum torque will be:

T = D2Hcu

2 + 2 2 δr-rcu

D/20 (3.11)

= D2Hcu

2 +

4πr3

3cu

0

D/2

= D2H2

1+ D3H

cu

For a vane blade where H = 2D: T = 3.667D3 cu (3.12)


3-18 March 2009

Figure 3.13 Assumed Geometry of Shear Surface for Conventional Interpretation of the Vane Test If it is assumed that the shear stress mobilized by the soil is linearly proportional to displacement, up to failure, then another simple assumption (Skempton 1948), that the shear stress on the top and bottom of the cylindrical shear surface has a triangular distribution, is sometimes adopted. For the rectangular vane this leads to the equation:

T = D2H2

1+ D4H

cu (3.13)

For a vane blade where H = 2D: T = 3.53D3cu (3.14) giving only 4% difference in shear strength from that obtained using the uniform assumption. Undrained Strength and Sensitivity The conventional interpretation for obtaining the undrained shear strength from the recorded maximum torque (T) assumes a uniform distribution of shear stresses both top and bottom along the blades and a vane with height-to-width ratio H/D = 2 (Chandler, 1988), as given in Eq. 3-11 above, regardless of units so long as torque T and width D are in consistent units (e.g., kN-m and meters, respectively, to provide vane strength cuv in kN/m2). The test is normally reserved for soft to stiff materials with cuv < 200 kPa. (2 tsf). After the peak cuv is obtained, the vane is rotated quickly through 10 complete revolutions and the remoulded (or "residual") value is recorded. The in-situ sensitivity of the soil is defined by:


March 2009 3-19

St = cu(peak)/cu(remolded) (3.15) For the commercial vanes in common use, the following expressions for vanes with blade heights that are twice their widths (H/D = 2) are obtained: Rectangular (iT = 0° and iB = 0°): suv = 0.273 Tmax/D3 (3.16a) Nilcon (iT = 0° and iB = 45°): suv = 0.265 Tmax/D3 (3.16b) Geonor (iT = 45° and iB = 45°): suv = 0.257 Tmax/D3 (3.16c) Vane Correction Factor It is very important that the measured vane strength be corrected prior to use in stability analyses involving embankments on soft ground, bearing capacity, and excavations in soft clays. The mobilized shear strength is given by:

τmobilized = μR suv (3.17)

where μR = empirical correction factor that has been related to plasticity index (PI) and/or liquid limit (LL) based on back-calculation from failure case history records of full-scale projects. An extensive review of the factors and relationships affecting vane measurements in clays and silts with PI > 5% recommends the following expression (Chandler, 1988):

μR = 1.05 - b (PI)0.5 (3.18) where the parameter b is a rate factor that depends upon the time-to-failure (tf in minutes) and given by: b = 0.015 + 0.0075 log tf (3.19) The combined relationships are shown in Fig. 3.14. For guidance, embankments on soft ground are normally associated with tf on the order of 104 minutes because of the time involved in construction using large equipment.


3-20 March 2009

Figure 3.14 Vane Correction Factor (:R) Expressed in Terms of Plasticity Index and Time to Failure. (Adapted from Chandler, 1988). Note: For Stability Analyses Involving Normal Rates of Embankment

Construction, the Correction Factor is Taken at the Curve Corresponding to Tf = 10,000 Minutes. It has been shown that the mobilized undrained shear strength back-calculated from failure case histories involving embankments, foundations, and excavations in soft clays are essentially independent of plasticity index (Terzaghi, et al. 1996). For further information, a detailed review of the device, the procedures, and methods of interpretation for the VST are given by Chandler (1988). 3.4 SUMMARY ON IN-SITU GEOTECHNICAL METHODS In-situ physical testing provide direct information concerning the subsurface conditions, geo-stratigraphy, and engineering properties prior to design, bids, and construction on the ground. In soils, in-situ geotechnical tests include penetration-type (Standard Penetration Test (SPT), Cone Penetration Test (CPT), Cone Penetrometer Test / Piezocone Test (CPTu), Flat Dilatometer Test (DMT), Cone Pressuremeter (CPMT), Vane Shear Test (VST)) and probing-type (Pressuremeter Test (PMT), Self-boring Pressurementer(SBP)) methods to directly obtain the response of the geomaterials under various loading situations and drainage conditions. The general applicability of the test method depends in part on the geo-material types encountered during the site investigation, as shown by Figure 3.15. The relevance of each test also depends on the project type and its requirements. Commonly used penetration type tests are Standard Penetration Test (SPT), Cone Penetration Test (CPT) and Vane Shear Test (VST). Other tests are carried out for special purposes and requirements.


March 2009 3-21

Figure 3.15 Relevance of In-Situ Tests to Different Soil Types

The evaluation of strength, deformation, flow, and time-rate behaviour of soil materials can be derived from selected tests or combinations of these test methods. Together, information from these tests allow for the rational and economical selection for deciding foundation types for bridges and buildings, safe embankment construction over soft ground, cut angles for adequate slope stability, and lateral support for underground excavations. 3.5 GROUNDWATER INVESTIGATIONS 3.5.1 General Groundwater conditions and the potential for groundwater seepage are fundamental factors in virtually all geotechnical analyses and design studies. Accordingly, the evaluation of groundwater conditions is a basic element of almost all geotechnical investigation programs. Groundwater investigations are of two types as follows:

o Determination of groundwater levels and pressures and o Measurement of the permeability of the subsurface materials.

Determination of groundwater levels and pressures includes measurements of the elevation of the groundwater surface or water table and its variation with the season of the year; the location of perched water tables; the location of aquifers (geological units which yield economically significant amounts of water to a well); and the presence of artesian pressures. Water levels and pressures may be measured in existing wells, in boreholes and in specially-installed observation wells. Piezometers are used where the measurement of the ground water pressures are specifically required (i.e. to determine excess hydrostatic pressures, or the progress of primary consolidation). Determination of the permeability of soil or rock strata is needed in connection with surface water and groundwater studies involving seepage through earth dams, yield of wells, infiltration, excavations and basements, construction dewatering, contaminant migration from hazardous waste

In-s

itu T

est

Met

hod

Grain size (mm)

SPT

CPT

DMT

PMT

VST

Geophysics


3-22 March 2009

spills, landfill assessment, and other problems involving flow. Permeability is determined by means of various types of seepage, pressure, pumping, and flow tests. 3.5.2 Determination of Ground Water Levels and Pressures Observations of the groundwater level and pressure are an important part of all geotechnical explorations, and the identification of groundwater conditions should receive the same level of care given to soil descriptions and samples. Measurements of water entry during drilling and measurements of the groundwater level at least once following drilling should be considered a minimum effort to obtain water level data, unless alternate methods, such as installation of observation wells, are defined by the geotechnical engineer. 3.5.3 Field Measurement of Permeability The permeability (k) is a measure of how easily water and other fluids are transmitted through the geo-material and thus represents a flow property. In addition to groundwater related issues, it is of particular concern in geo-environmental problems. The parameter k is closely related to the coefficient of consolidation (cv) since time rate of settlement is controlled by the permeability. In geotechnical engineering, we designate small k = coefficient of permeability or hydraulic conductivity (units of cm/sec), which follows Darcy's law: q = kiA (3.20) where q = flow (cm3/sec), i = hydraulic gradient, and A = cross-sectional area of flow. Laboratory permeability tests may be conducted on undisturbed samples of natural soils or rocks, or on reconstituted specimens of soil that will be used as controlled fill in embankments and earthen dams. Field permeability tests may be conducted on natural soils (and rocks) by a number of methods, including simple falling head, packer (pressurized tests), pumping (drawdown), slug tests (dynamic impulse), and dissipation tests. A brief listing of the field permeability methods is given in Table 3.5. Field permeability arrangement for soil and rock are presented in Figure 3.16 and Figure 3.17.

March 200

Variou

PumpiDoubleinfiltroInfiltrosealedVariouSlug teHydra

Consta

Pressutechni

Piezoc

Dilatom

Falling

09

Test Method

us Field Metho

ing tests e-ring ometer ometer with d ring us field methoests ulic fracturing

ant head injec

ure pulse que

cone dissipatio

meter dissipa

g head tests

Chap

Table 3.5 Fie

A

ods Soil a

Draw

Surfa

Surfa

ods Soils Soils

g Rock

ction Low-rocksLow-rocks

on Low t

ation Low t

Cased

Figure 3.16

pter 3 IN-SITU

eld Methods f

Applicable Soil

and Rock Aqu

wdown in soils

ace fill soils

ace soils

in vadose zonat depth in-situ permeability s permeability s

to medium k

to medium k

d borehole in

Field Permea

GEOTECHNICA

for Measurem

ls

ifers A

s A

A

A

ne AAA

A

A

soils Ho

soils Ro

soils Lam

ability Test Arr

AL TESTING

ment of Perme

Re

ASTM D4043

ASTM D4050

ASTM D3385

ASTM D5093

ASTM D5126 ASTM D4044 ASTM D4645

ASTM D4630

ASTM D4630

oulsby & The (1988)

bertson et al.(1988)

mbe & Whitma(1979)

rangement fo

eability

eference

.

an BS-59

or Soil

3-

30 –(1988)

-23


3-24 March 2009

REFERENCES [1] American Association of State Highway and Transportation Officials (AASHTO) (1988). Manual on Subsurface Investigations, Developed by the Subcommittee on Materials, Washington, D.C. [2] American Society for Testing & Materials. (2000). ASTM Book of Standards, Vol. 4, Section 08 and Baguelin, F., Jezequel, J. F., and Shields, D. H. (1978). The Pressuremeter and Foundation Engineering, Trans Tech Publication, Switzerland. [3] Baldi, G., Bellotti, R., Ghionna, V., Jamiolkowski, M. and LoPresti, D.C. (1989). "Modulus of sands from CPTs and DMTs", Proceedings, 12th International Conference on Soil Mechanics & Foundation Engineering, Vol. 1, Rio de Janeiro, 165-170. [4] Briaud, J. L. (1989). “The pressuremeter test for highway applications.” Report FHWA-IP-89-008, Federal Highway Administration, Washington, D.C., 148. [5] Burns, S.E. and Mayne, P.W. (1996). “Small- and high-strain measurements of in-situ soil properties using the seismic cone”. Transportation Research Record 1548, Natl. Acad. Press, Wash., D.C., 81-88. [6] Burns, S.E. and Mayne, P.W. (1998). “Monotonic and dilatory pore pressure decay during piezocone tests”. Canadian Geotechnical Journal, Vol. 35 (6), 1063-1073. [7] Campanella, R.G. (1994). "Field methods for dynamic geotechnical testing", Dynamic Geotechnical Testing II (STP 1214), ASTM, Philadelphia, 3-23. [8] Campanella, R. G., and Robertson, P. K. (1981). “Applied cone research”, Cone Penetration Testing and Experience, ASCE, Reston/VA, 343-362. [9] Chandler, R.J. (1988). “The in-situ measurement of the undrained shear strength of clays using the field vane”. Vane Shear Strength Testing in Soils: Field and Laboratory Studies. ASTM STP 1014, American Society for Testing & Materials, West Conshohocken/PA, 13-44. [10] Chen, B.S-Y. and Mayne, P.W. (1996). “Statistical relationships between piezocone measurements and stress history of clays”. Canadian Geotechnical Journal, Vol. 33 (3), 488-498. [11] Driscoll, F. G. (1986). Groundwater and Wells, 2nd Edition, Johnson Filtration Systems, Inc., St. Paul, MN, 1089 p. [12] Dunnicliff, J. (1988). Geotechnical Instrumentation for Monitoring Field Performance, John Wiley & Sons, Inc., New York. [13] Fahey, M. and Carter, J.P. (1993). “A finite element study of the pressuremeter in sand using a nonlinear elastic plastic model”, Canadian Geotechnical Journal, Vol. 30 (2), 348-362. [14] Finn, P. S., Nisbet, R. M., and Hawkins, P. G. (1984). "Guidance on pressuremeter, flat dilatometer and cone penetration tests in sand." Géotechnique, Vol. 34 (1), 81-97. [15] Greenhouse, J.P., Slaine, D.D., and Gudjurgis, P. (1998). Application of Geophysics in Environmental Investigations, Matrix Multimedia Publishing, Toronto. Hatanaka, M. and Uchida, A. (1996).”


March 2009 3-25

[16] “Empirical correlation between penetration resistance and effective friction of sandy soil”, Soils & Foundations, Vol. 36 (4), Japanese Geotechnical Society, 1-9. [17] Hegazy, Y.A. (1998). Delineating geostratigraphy by cluster analysis of piezocone data. PhD Thesis, School of Civil and Environmental Engineering, Georgia Institute of Technology, Atlanta, 464 p. [18] Hoar, R.J. and Stokoe, K.H. (1978), "Generation and measurement of shear waves in-situ", Dynamic Geotechnical Testing (STP 654), ASTM, Philadelphia, 3-29. [19] Holtz, W. G., and Gibbs, H. J. (1979). Discussion of “SPT and relative density in coarse sand.” Journal of Geotechnical Engineering, ASCE, Vol. 105 (3), 439-441. [20] Houlsby, G.T. and Teh, C.I. (1988). “Analysis of the piezocone in clay”, Penetration Testing 1988, Vol. 2, Balkema, Rotterdam, 777-783. [21] Jamiolkowski, M., Ladd, C. C., Germaine, J. T., and Lancellotta, R. (1985). “New developments in field and laboratory testing of soils.” Proceedings, 11th International Conference on Soil Mechanics & Foundation Engineering, Vol. 1, San Francisco, 57-153. [22] Kovacs, W.D., Salomone, L.A., and Yokel, F.Y. (1983). “Energy Measurements in the Standard Penetration Test.” Building Science Series 135, National Bureau of Standards, Washington, 73. [23] Kulhawy, F.H. and Mayne, P.W. (1991). Relative density, SPT, and CPT interrelationships. Calibration Chamber Testing, (Proceedings, ISOCCT, Potsdam), Elsevier, New York, 197-211. [24] Kulhawy, F.H., Trautmann, C.H., and O’Rourke, T.D. (1991). “The soil-rock boundary: What is it and where is it?”. Detection of and Construction at the Soil/Rock Interface, GSP No. 28, ASCE, Reston/VA, 1-15. [25] Kulhawy, F.H. and Phoon, K.K. (1993). “Drilled shaft side resistance in clay soil to rock”, Design and Performance of Deep Foundations: Piles & Piers in Soil & Soft Rock, GSP No. 38, ASCE, Reston/VA, 172-183. [26] Leroueil, S. and Jamiolkowski, M. (1991). “Exploration of soft soil and determination of design parameters”, Proceedings, GeoCoast’91, Vol. 2, Port & Harbor Res. Inst., Yokohama, 969-998. [27] Lunne, T., Powell, J.J.M., Hauge, E.A., Mokkelbost, K.H., and Uglow, I.M. (1990). “Correlation for dilatometer readings with lateral stress in clays”, Transportation Research Record 1278, National Academy Press, Washington, D.C., 183-193. [28] Lunne, T., Lacasse, S., and Rad, N.S. (1994). “General report: SPT, CPT, PMT, and recent developments in in-situ testing”. Proceedings, 12th International Conference on Soil Mechanics & Foundation Engineering, Vol. 4, Rio de Janeiro, 2339-2403. [29] Lunne, T., Robertson, P.K., and Powell, J.J.M. (1997). Cone Penetration Testing in Geotechnical Practice, Blackie-Academic Publishing/London, EF SPON Publishing, U.K., 317 p. [30] Mair, R. J., and Wood, D. M. (1987). "Pressuremeter testing methods and interpretation." Ground Engineering Report: In-Situ Testing,(CIRIA), Butterworths, London, U.K.


3-26 March 2009

[31] Marchetti, S. (1980). “In-situ tests by flat dilatometer”, Journal of the Geotechnical Engineering Division (ASCE), Vol. 107 (3), 832-837. [32] Marchetti, S. (1997). "The flat dilatometer: design applications", Proceedings, Third International Geotechnical Engineering Conference, Cairo University, Egypt, 1-25. [33] Marcuson, W.F. and Bieganousky, W.A. (1977). "SPT and relative density in coarse sands", Journal of the Geotechnical Engineering Division (ASCE), Vol. 103 (GT11), 1295-1309. [34] Mayne, P. W., and Mitchell, J. K. (1988). "Profiling of overconsolidation ratio in clays by field vane." Canadian Geotechnical Journal, Vol. 25 (1), 150-158. [35] Mayne, P.W., Kulhawy, F.H., and Kay, J.N. (1990). “Observations on the development of pore water pressures during piezocone tests in clays”. Canadian Geotechnical Journal, Vol. 27 (4), 418-428. [36] Mayne, P.W. and Kulhawy, F.H. (1990). “Direct & indirect determinations of in-situ K0 in clays.” Transportation Research Record 1278, National Academy Press, Washington, D.C., 141-149. [37] Mayne, P.W. (1991). “Determination of OCR in clays by piezocone tests using cavity expansion and Mayne, P.W. and Rix, G.J. (1993). "Gmax-qc relationships for clays", ASTM Geotechnical Testing Journal, Vol. 16 (1), 54-60. [38] critical state concepts.” Soils and Foundations, Vol. 31 (2), 65-76. [39] Mayne, P.W., Mitchell, J.K., Auxt, J., and Yilmaz, R. (1995). “U.S. national report on the CPT”. Proceedings, International Symposium on Cone Penetration Testing (CPT’95), Vol. 1, Swedish Geotechnical Society, Linköping, 263-276. [40] Mayne, P.W. (1995). “Profiling yield stresses in clays by in-situ tests”. Transportation Research Record 1479, National Academy Press, Washington, D.C., 43-50. [41] Mayne, P.W. (1995). “CPT determination of OCR and Ko in clean quartz sands”. Proceedings, CPT’95, Vol. 2, Swedish Geotechnical Society, Linkoping, 215-220. [42] Mayne, P.W., Robertson, P.K., and Lunne, T. (1998). “Clay stress history evaluated from seismic piezocone tests”. Geotechnical Site Characterization, Vol. 2, Balkema, Rotterdam, 1113-1118. [43] Mayne, P.W. and Martin, G.K. (1998). “Commentary on Marchetti flat dilatometer correlations in soils.” ASTM Geotechnical Testing Journal, Vol. 21 (3), 222-239. [44] Mayne, P.W., Schneider, J.A., and Martin, G.K. (1999). "Small- and large-strain soil properties from seismic flat dilatometer tests", Pre-Failure Deformation Characteristics of Geomaterials, Vol. 1 (Torino), Balkema, Rotterdam, 419-426. [45] Mayne, P.W. (2001). “Stress-strain-strength-flow parameters from enhanced in-situ tests”, Proceedings, International Conference on In-Situ Measurement of Soil Properties & Case Histories (In-Situ 2001), Bali, Indonesia, 47-69. [46] Parez, L. and Faureil, R. (1988). “Le piézocône. Améliorations apportées à la reconnaissance de sols”. Revue Française de Géotech, Vol. 44, 13-27.


March 2009 3-27

[47] Rehm, B.W., Stolzenburg, T. R., and Nichols, D. G. (1985). “Field measurement methods for hydrogeologic investigations: a critical review of the literature.” EPRI Report No. EA-4301, Electric Power Research Institute, Palo Alto, CA. [48] Robertson, P.K. and Campanella, R.G. (1983). “Interpretation of cone penetration tests: Part I - sands; Part II - clays”. Canadian Geotechnical Journal, Vol. 20 (4), 719-745. [49] Robertson, P.K., Campanella, R.G., and Wightman, A. (1983). “SPT-CPT correlations”, Journal of the Geotechnical Engineering Division (ASCE), Vol. 109 (11), 1449-1459. [50] Robertson, P.K. (1986). “In-situ testing and its application to foundation engineering”, Canadian Geotechnical Journal, Vol. 23 (4), 573-584. [51] Robertson, P.K., Campanella, R.G., Gillespie, D., and Grieg, J. (1986). “Use of piezometer cone data”. Use of In-Situ Tests in Geotechnical Engineering, GSP No. 6, ASCE, New York, 1263-1280. [52] Robertson, P.K., Campanella, R.G., Gillespie, D., and Rice, A. (1986). “Seismic CPT to measure in-situ shear wave velocity”. Journal of Geotechnical Engineering 112 (8), 71-803. [53] Robertson, P.K., Campanella, R.G., Gillespie, D. and By, T. (1988). “Excess pore pressures and the flat dilatometer test”, Penetration Testing 1988, Vol. 1, Balkema, Rotterdam, 567-576. [54] Robertson, P.K. (1990). “Soil classification using the cone penetration test”. Canadian Geotechnical Journal, Vol. 27 (1), 151-158. [55] Santamarina, J.C., Klein, K. and Fam, M.A. (2001). Soils and Waves, Particulate Materials Behavior, Characterization, & Process Monitoring, John Wiley & Sons, Ltd., New York, 488 p. [56] Schmertmann, J.H. (1986). “Suggested method for performing the flat dilatometer test”, ASTM Geotechnical Testing Journal, Vol. 9 (2), 93-101. [57] Skempton, A.W. (1986). “SPT procedures and the effects in sands of overburden pressure, relative density, particle size, aging, and overconsolidation”. Geotechnique, Vol. 36, No. 3, 425-447. [58] Stokoe, K. H., and Woods, R. D. (1972). "In-situ shear wave velocity by cross-hole method." Journal of the. Soil Mechanics &.Foundations Division, ASCE, 98 (5), 443-460. [59] Stokoe, K. H., and Hoar, R. J. (1978). "Variables affecting in-situ seismic measurement." Proceedings, Earthquake Engineering and Soil Dynamics, ASCE, Pasadena, Ca, 919-938. [60] Tanaka, H. and Tanaka, M. (1998). "Characterization of sandy soils using CPT and DMT", Soils and Foundations, Vol. 38 (3), 55-67 [61] Tatsuoka, F. and Shibuya, S. (1992). “Deformation characteristics of soils & rocks from field & lab tests.” Report of the Institute of Industrial Science 37 (1), Serial No. 235, University of Tokyo, 136 p. [62] Tavenas, F., LeBlond, P., Jean, P., and Leroueil, S. (1983). “The permeability of natural soft clays: Parts I and II”, Canadian Geotechnical Journal, Vol. 20 (4), 629-660.


3-28 March 2009

[63] Teh, C.I. and Houlsby, G.T. (1991). “An analytical study of the cone penetration test in clay”. Geotechnique, Vol. 41 (1), 17-34. [64] U.S. Department of the Interior, Bureau of Reclamation. (1973). Design of small dams, United States Government Printing Office, Washington, D.C. [65] U.S. Army Corps of Engineers. (1951). "Time lag and soil permeability in groundwater observations." Waterways Experiment Station, Bulletin No. 36, Vicksburg, MS. [66] U.S. Department of the Interior, Bureau of Reclamation (1960). Earth manual, United States Government Printing Office, Washington, D.C. [67] Windle, D., and Wroth, C. P. (1977). "In-situ measurement of the properties of stiff clays." Proceedings, 9th International Conference on Soil Mechanics and Foundation Engineering, Vol. 1, Tokyo, Japan, 347-352.

CHAPTER 4 LAB TESTING FOR SOILS

Chapter 4 LABORATORY TESTING FOR SOILS

March 2009 4-i

z Table of Contents


List of Table ........................................................................................................................... 4-ii


4.1 GENERAL .................................................................................................................... 4-1

4.2 WEIGHT – VOLUME CONCEPTS .................................................................................... 4-1

4.3 LOAD-DEFORMATION PROCESS IN SOILS ..................................................................... 4-2

4.4 PRINCIPLES OF EFFECTIVE STRESS .............................................................................. 4-3

4.5 OVERBURDEN STRESS ................................................................................................. 4-3

4.6 TESTS FOR GEOTECHNICAL PARAMETERS .................................................................... 4-4

4.6.1 Classification Tests ........................................................................................ 4-5

4.6.2 Chemical and Electro-chemical Tests .............................................................. 4-7

4.6.3 Compaction Related Tests .............................................................................. 4-8

4.6.4 Compressibility, Permeability and Durability Tests ............................................ 4-9

4.6.5 Consolidation and Permeability Tests in Hydraulic Cells and with Pore Pressure Measurement ................................................................................................ 4-9

4.6.6 Shear Strength Tests (Total and Effective Stresses) ........................................ 4-10

REFERENCES ....................................................................................................................... 4-16


4-ii March 2009

List of Table


4.1 Terms in Weight – Volume Relations (After Cheney And Chassie, 1993) 4-1

4.2 Unit Weight – Volume Relationships 4-2

4.3 Available Chemical Tests 4-7

List of Figures


4.1 Typical Particle Size Distribution 4-5

4.2 Casagrande Plot Showing Classification of Soil into Groups 4-7

4.3 Typical Compaction Curves 4-8

4.4 Consolidation Test Apparatus 4-10

4.5 Bishop Direct Shear Box 4-12

4.6 Triaxial Cell 4-13


March 2009 4-1

4 LABORATORY TESTING FOR SOILS

4.1 GENERAL Laboratory testing of soils is a fundamental element of geotechnical engineering. The complexity of testing required for a particular project may range from a simple moisture content determination to specialized strength and stiffness testing. Since testing can be expensive and time consuming, the geotechnical engineer should recognize the projects issues ahead of time so as to optimize the testing program, particularly strength and consolidation testing. Before describing the various soil test methods, soil behavioural under load will be examined and common soil mechanics terms introduced. The following discussion includes only basic concepts of soil behaviour. However, the engineer must grasp these concepts in order to select the appropriate tests to model the in-situ conditions. The terms and symbols shown will be used in all the remaining modules of the course. Basic soil mechanics textbooks should be consulted for further explanation of these and other terms. 4.2 WEIGHT – VOLUME CONCEPTS A sample of soil is usually composed of soil grains, water and air. The soil grains are irregularly shaped solids which are in contact with other adjacent soil grains. The weight and volume of a soil sample depends on the specific gravity of the soil grains (solids), the size of the space between soil grains (voids and pores) and the amount of void space filled with water. Common terms associated with weight-volume relationships are shown in Table 4.1.

Table 4.1 Terms in Weight – Volume Relations (After Cheney And Chassie, 1993)

Property Symbol Units1 How obtained (AASHTO/ASTM/BSS) Direct Applications

Moisture Content w D By measurement (T 265/D 4959/BS1377-Part 2)

Classification and in weight-volume relations

Specific Gravity Gc D By measurement (T 100/D 854 BS1377-Part 2) Volume computations

Unit weight FL-3 By measurement or from weight-volume relations

Classification and for pressure computations

Porosity n D From weight-volume relations Defines relative volume of solids to total volume of soil

Void Ratio e D From weight-volume relations Defines relative volume of solids to total volume of soil

1 F = Force or weight; L = Length; D = Dimensionless. Although by definition, moisture content is a dimensionless fraction (ratio of weight of water of solids), it is commonly reported in percent by multiplying the fraction by 100.

Of particular note is the void ratio (e) which is a general indicator of the relative strength and compressibility of a soil sample, i.e., low void ratios generally indicates strong soils of low compressibility, while high void ratios are often indicative of weak and highly compressible soils. Selected weight-volume (unit weight) relations are presented in Table 4.2.


4-2 March 2009

Table 4.2 Unit Weight – Volume Relationships

Case Relationship Applicable Geomaterials

Soil Identities 1. Gδ w = S e 2. Total Unit Weight:

= 1+w1+e

Gs w

All types of soils and rocks

Limiting Unit Weight Solid phase only: w=e=0:

rock = Gs w

Maximum expected value for solid silica is 27 kN/m3

Dry Unit Weight For w=0 (all air in void space):

d = Gs w/(1+e)

Use for clean sands and dry soils above groundwater table

Moist Unit Weight (Total Unit Weight)

Variable amounts of air and water:

t = Gs w(1+w)/(1+e) with e = Gδ w/S

Partially-saturated soils above water table; depends on degree of saturation (S, as decimal)

Saturated Unit Weight Set S = 1 (all voids with water):

sat = w(Gs+e)/(1+e)

All soils below water table; Saturated clays and silts above water table with full capillarity

Hierarchy d ≤ t ≤ sat < rock

Check on relative values

Note: w = 9.8 kN/m3 (62.4 pcf) for fresh water 4.3 LOAD-DEFORMATION PROCESS IN SOILS When a load is applied to a soil sample, the deformation which occurs will depend on the grain-to-grain contact (inter-granular) forces and the amount of water in the voids. If no porewater exists, the sample deformation will be due to sliding between soil grains and deformation of the individual soil grains. The rearrangement of soil grains due to sliding accounts for most of the deformation. Adequate deformation is required to increase the grain contact areas to take the applied load. As the amount of pore water in the void increases, the pressure it exerts on soil grains will increase and reduce the inter-granular contact forces. In fact, tiny clay particles may be forced completely apart by water in the pore space. Deformation of a saturated soil is more complicated than that of dry soil as water molecules, which fill the voids, must be squeezed out of the sample before readjustment of soil grains can occur. The more permeable a soil is, the faster the deformation under load will occur. However, when the load on a saturated soil is quickly increased, the increase is carried entirely by the pore water until drainage begins. Then more and more load is gradually transferred to the soil grains until the excess pore pressure has dissipated and the soil grains readjust to a denser configuration. This process is called consolidation and results in a higher unit weight and a decreased void ratio.


March 2009 4-3

4.4 PRINCIPLES OF EFFECTIVE STRESS The consolidation process demonstrates the very important principle of effective stress, which will be used in all the remaining modules of this course. Under an applied load, the total stress in a saturated soil sample is composed of the inter-granular stress and porewater pressure (neutral stress). As the porewater has zero shear strength and is considered incompressible, only the inter-granular stress is effective in resisting shear or limiting compression of the soil sample. Therefore, the inter-granular contact stress is called the effective stress. Simply stated, this fundamental principle states that the effective stress (σ’) on any plane within a soil mass is the net difference between the total stress (σt) and porewater pressure (u). When pore water drains from soil during consolidation, the area of contact between soil grains increases, which increases the level of effective stress and therefore the soil’s shear strength. In practice, staged construction of embankments is used to permit increase of effective stress in the foundation soil before subsequent fill load is added. In such operations the effective stress increase is frequently monitored with piezometers to ensure the next stage of embankment can be safely placed. Soil deposits below the water table will be considered saturated and the ambient pore pressure at any depth may be computed by multiplying the unit weight of water (γw) by the height of water above that depth. For partially saturated soil, the effective stress will be influenced by the soil structure and degree of saturation (Bishop, et. al., 1960). In many cases involving silts and clays, the continuous void spaces that exist in the soil behave as capillary tubes of variable cross-section. Due to capillarity, water may rise above the static groundwater table (phreatic surface) as a negative porewater pressure and the soils may be nearly or fully saturated. 4.5 OVERBURDEN STRESS The purpose of laboratory testing is to simulate in-situ soil loading under controlled boundary conditions. Soils existing at a depth below the ground surface are affected by the weight of the soil above that depth. The influence of this weight, known generally as the overburden stress, causes a state of stress to exist which is unique at that depth for that soil. When a soil sample is removed from the ground, that state of stress is relieved as all confinement of the sample has been removed. In testing, it is important to re-establish the in-situ stress conditions and to study changes in soil properties when additional stresses representing the expected design loading are applied. In this regard, the effective stress (grain-to-grain contact) is the controlling factor in shear, state of stress, consolidation, stiffness, and flow. Therefore, the designer should try to re-establish the effective stress condition during most testing. The test confining stresses are estimated from the total, hydrostatic, and effective overburden stresses. The engineer’s first task is determining these stress and pressure variations with depth. This involves determining the total unit weights (density) for each soil layer in the subsurface profile, and determining the depth of the water table. Unit weight may be accurately determined from density tests on undisturbed samples or estimated from in-situ test measurements. The water table is routinely recorded on the boring logs, or can be measured in open standpipes, piezometers, and dissipation tests during CPTs and DMTs. The total vertical (overburden) stress (σvo) at any depth (z) may be found as the accumulation of total unit weights ( t) of the soil strata above that depth: σvo = t dz= ∑ t ∆z (4.1)


4-4 March 2009

For soils above the phreatic surface, the applicable value of total unit weight may be dry, moist, or saturated depending upon the soil type and degree of capillarity (see Table 4.2). For soil elements situated below the groundwater table, the saturated unit weight is normally adopted. The hydrostatic pressure depends upon the degree of saturation and level of the phreatic surface and is determined as follow: Soil elements above water table: uo = 0 (Completely dry) (4.2a) = w (z-zw) (Full capillarity) (4.2b) Soils elements below water table: uo = w (z-zw) (4.2c) where z = depth of soil element, zw = depth to groundwater table. Another case involves partial saturation with intermediate values between (4.2a and 4.2b) which literally vary daily with the weather and can be obtained via tensiometer measurements in the field. Usual practical calculations adopt (4.2a) for many soils, yet the negative capillary values from (4.2b) often apply to saturated clay and silt deposits. The effective vertical stress is obtained as the difference between (4.1) and (4.2): σvo’ = σvo - uo (4.3) A plot of effective overburden profile with depth is called a ’v diagram and is extensively used in all aspects of foundation testing and analysis (see Holtz & Kovacs, 1981; Lambe & Whitman, 1979). 4.6 TESTS FOR GEOTECHNICAL PARAMETERS A wide range of tests has been used to determine the geotechnical parameters required in calculations for example, of bearing capacity, slope stability, earth pressure and settlement. Geotechnical calculations remain almost entirely semi-empirical in nature; it has been said that when calculating the stability of a slope one uses the ‘wrong’ slip circle with the ‘wrong’ shear strength to arrive at a satisfactory answer. For this reason testing requirements differ considerably from region to region. The new British Standard (BS 1377:1990.) is divided into nine separate parts:

Part 1 General requirements and sample preparation Part 2 Classification tests Part 3 Chemical and electro-chemical tests Part 4 Compaction-related tests Part 5 Compressibility, permeability and durability tests Part 6 Consolidation and permeability tests in hydraulic cells and with pore pressure

measurement Part 7 Shear strength tests (total stress) Part 8 Shear strength tests (effective stress) Part 9 In situ tests.

March 200

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4-6 March 2009

7. Spread soil in a tray and cover with water and sodium hexametaphosphate (2 g/l). 8. Stir frequently for 1 h, to break down and separate clay particles. 9. Place soil in small batches on a 2mm sieve resting on a 63 m sieve and wash gently to remove

fines. 10. When clean, place the material retained in an oven and dry at 105—110°C. 11. Sieve through standard mesh sizes between 20mm and 6.3 mm using the dry sieving

procedure. Note weights retained on each sieve. 12. f more than 150 g passes the 6.3mm mesh, split the sample by riffling to give 100—150g. 13. Sieve through standard mesh sizes between 5mm and 63 tm sieve. The particle size distribution of the fine soil fraction, between about 0.1 mm and 1 µm may be determined by one of two British Standard sedimentation tests (BS 1377:part 2, clauses 9.4, 9.5). Soil is sedimented through water, and Stokes’ law, which relates the terminal velocity of a spherical particle falling through a liquid of known viscosity to its diameter and specific gravity, is used to deduce the particle size distribution. Sedimentation tests make a number of important assumptions. Since Stokes’ law is used, the following assumptions are implied (Allen 1975). 1. The drag force on each particle is due entirely to viscous forces within the fluid. The particles

must be spherical, smooth and rigid, and there must be no slippage between them and the fluid.

2. Each particle must move as if it were a single particle in a fluid of infinite extent. 3. The terminal velocity must be reached very shortly after the test starts. 4. The settling velocity must be slow enough so that inertia effects are negligible. 5. The fluid must be homogeneous compared with the size of the particle. Plasticity tests The plasticity of soils is determined by using relatively simple remoulded strength tests. The plastic limit is the moisture content of the soil under test when remoulded and rolled between the tips of the fingers and a glass plate such that longitudinal and transverse cracks appear at a rolled diameter of 3 mm. At this point the soil has a stiff consistency. The liquid limit of a soil can be determined using the cone penetrometer or the Casagrande apparatus (BS 1377:1990:part 2, clauses 4.3, 4.5 / ASTM D-423-54T and ASTM D-424-54T). One of the major changes introduced by the 1975 British Standard (BS 1377 ) was that the preferred method of liquid limit testing became the cone penetrometer. This preference is reinforced in the revised 1990 British Standard which refers to the cone penetrometer as the ‘definitive method’. The cone penetrometer is considered a more satisfactory method than the alternative because it is essentially a static test which relies on the shear strength of the soil, whereas the alternative Casagrande cup method introduces dynamic effects. In the penetrometer test, the liquid limit of the soil is the moisture content at which an 80 g, 300 cone sinks exactly 20 mm into a cup of remoulded soil in a 5s period. Plasticity tests are widely used for classification of soils (Fig. 4.2) into groups on the basis of their position on the Casagrande chart (Casagrande 1948), but in addition they are used to determine the suitability of wet cohesive fill for use in earthworks, and to determine the thickness of sub-base required beneath highway pavements (Road Research Laboratory 1970).

March 200

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March 20

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March 2009 4-9

4.6.4 Compressibility, Permeability and Durability Tests Laboratory determinations of the permeability of granular soils can be made using the constant head and falling head permeameter tests (BS 1377: part 5:1990, clause 5). For granular soils any values of permeability must be regarded as approximate, since several important factors affect the accuracy of these tests. Cohesive soils can be tested for coefficient of permeability in the laboratory, and indeed it was for this purpose that Terzaghi (1923) produced the one-dimensional consolidation theory. Terzaghi noted that smear on the specimen boundaries greatly affected the measured soil permeability in his permeameter tests, and used an oedometer test in order that all water flow would occur out of the sample. Thus the coefficient of permeability can be obtained from triaxial or hydraulic consolidation tests since: k = cvmv w (4.4) where k = coefficient of permeability, cv = coefficient of consolidation, mv = the coefficient of compressibility, and w = density of water. 4.6.5 Consolidation and Permeability Tests in Hydraulic Cells and with Pore

Pressure Measurement Consolidation tests are frequently required either to assess the amount of volume change to be expected of a soil under load, for example beneath a foundation, or to allow prediction of the time that consolidation will take. The effect of predictions based on consolidation test results can be very serious, for example leading to the use of piling beneath structures, and the use of sand drains or stage construction for embankments. It is therefore important to appreciate the limitations of the commonly available test techniques. Three pieces of apparatus are in common use for consolidation testing. These are: a. The oedometer (Terzaghi 1923; Casagrande 1936); b. The triaxial apparatus (Bishop and Henkel 1962); and c. The hydraulic consolidation cell (Rowe and Barden 1966). a. Casagrande oedometer test The Casagrande oedometer test is most widely used. BS 1377: part 5:1990, clause 3 gives a standard procedure for the test. In this procedure the specimen is subjected to a series of pre-selected vertical stresses (e.g. 6, 12, 25, 50, 100, 200, 400, 800, 1600, 3200 kN/m2) each of which is held constant while dial gauge measurements of vertical deformation of the top of the specimen are made, and until movements cease (normally 24 h). b. Triaxial Dissipation Test The measurement of consolidation characteristics can be carried out in the triaxial dissipation test (Fig 4.6). The most common size of specimen is 102mm high x 102mm dia., and the test is carried out in a triaxial chamber such as might be used for a consolidated undrained triaxial compression test with pore pressure measurement. The specimen is compressed under the isotropic effective stress produced by the difference between the cell pressure and the back pressure, and volume change is recorded as a function of time, as in the consolidation stage of an effective strength triaxial compression test, but in addition pore pressure is measured at the base of the specimen. Drainage occurs upwards in the vertical direction but soil compression is three-dimensional, and for this reason the results of this test are not strictly comparable with those of an oedometer test. The compressibility determined from volume changes during the triaxial dissipation is greater than that


4-10 March 2009

measured under conditions of zero lateral strain, and the difference is most pronounced for overconsolidated clays and compacted soils. c. Hydraulic Consolidation Cell (Rowe Cells Consolidation Test) The conventional oedometer enables one to determine the consolidation characteristics in the vertical direction only. With some modifications, the hydraulic consolidation cell (Rowe cell) with radial drainage can measure the horizontal consolidation properties. The Rowe cell is an incremental loading test similar to a conventional oedometer test with a reasonably long testing duration. These cells, in which load is applied to the sample hydraulically, offer many advantages and considerably widen the scope of laboratory testing. In addition, the hydraulic loading system gives accurate control of applied loads over a wide range, including high pressures on large diameter samples.

(a) Schematic Diagram of Oedometer (b) Hydraulic Consolidation Cell

Figure 4.4 Consolidation Test Apparatus 4.6.6 Shear Strength Tests (Total and Effective Stresses) The principal tools available for strength determination include the California Bearing Ratio (CBR) apparatus, the Franklin Point Load Test apparatus (Franklin et al. 1971; Broch and Franklin 1972), the laboratory vane apparatus and various forms of direct shear and triaxial apparatus. For the purpose of relevance and application to DID related works, only the vane apparatus and the direct shear and triaxial tests are presented herein. Laboratory vane test The principles involved in the vane test are discussed in Section 3.3. Whilst the field vane typically uses a blade with a height of about 150 mm, the laboratory vane is a small-scale device with a blade height and width of about 12.7mm. The small size of the laboratory vane makes the device unsuitable for testing samples with fissuring or fabric, and therefore it is not very frequently used. The laboratory vane test is described in BS 1377: part 7:1990, clause 3. Direct shear test The vane apparatus induces shear along a more or less predetermined shear surface. In this respect the direct shear test carried out in the shear box apparatus (Skempton and Bishop 1950) is similar. Fig. 4.5 shows the basic components of the direct shear apparatus; soil is cut to fit tightly into a box which may be rectangular or circular in plan (Akroyd 1964; Vickers 1978; ASTM Part 19; Head 1982; BS 1377:1990), and is normally rectangular in elevation. The box is constructed to allow displacement along its horizontal mid-plane, and the upper surface of the soil is confined by a loading platen through which normal stress may be applied. Shear load is applied to the lower half of


March 2009 4-11

the box, the upper half being restrained by a proving ring or load cell which is used to record the shear load. The sample is not sealed in the shear box; it is free to drain from its top and bottom surfaces at all times. The cross-sectional area over which the specimen is sheared is assumed to remain constant during the test. The direct shear test has been used to carry out undrained and drained shear tests, and to determine residual strength parameters. Morgenstern and Tchalenko (1967) reported the results of optical measurements on clays at various stages during the direct shear test, and it is clear that at peak shear stress and beyond, failure structures (Reidels and thrust structures) are not coincident with the supposed imposed horizontal plane of failure. In addition, the restraints of the ends of the box create an even more markedly non-uniform shear surface. Since the direction of the failure planes, the magnitude and directions of principal stresses and the pore pressure are not determinable in a normal shear box experiment, its results are open to various interpretations (Hill 1950), and this test is now rarely used to determine undrained or peak effective strength parameters. Triaxial tests may be performed more conveniently and with better control.

4-12

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March 20

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March 2009 4-13

cap is produced by the cell fluid pressure and the ram force. The use of an aspect ratio of two ensures that the effects of the radial shear stresses between soil, and top cap and base-pedestal are insignificant at the centre of the specimen. The triaxial apparatus requires one or two self-compensating constant pressure systems, a volume change measuring device and several water pressure sensing devices. The ram force may be measured outside the cell using a proving ring, but most modern systems now use an internal electrical load cell mounted on the bottom of the ram. The ram is driven into the triaxial cell by an electrical loading frame which will typically have a capacity of 5000 or 10000 kgf and is capable of running at a wide range of constant speeds; triaxial tests are normally carried out at a controlled rate of strain increase.

Figure 4.6 Triaxial Cell The three most common forms of test are: a. The unconsoldiated undrained triaxial compression test, without pore water pressure

measurement (BS 1377:part 7:1990. clause 8); b. The consolidated undrained triaxial compression test, with pore water pressure measurement

(BS 1377:part 8:1990, clause 7); and c. The consolidated drained triaxial compression test, with volume change measurement (BS

1377:part 8:1990, clause 8). The unconsolidated undrained triaxial compression test is carried out on ‘undisturbed’ samples of clay in order to determine the undrained shear strength of the deposit in situ. Pore pressures are not measured during this test and therefore the results can only be interpreted in terms of total stress.


4-14 March 2009

Peak effective strength parameters (c' and φ') may be determined either from the results of consolidated undrained triaxial compression tests with pore pressure measurement or from consolidated drained triaxial compression tests. The former test is normally preferred because it can be performed more quickly and therefore more economically. The consolidated undrained triaxial compression test is normally performed in several stages, involving the successive saturation, consolidation and shearing of each of three specimens. Saturation is carried out in order to ensure that the pore fluid in the specimen does not contain free air. If this occurs, the pore air pressure and pore water pressure will differ owing to surface tension effects: the average pore pressure cannot be found as it will not be known whether the measured pore pressure is due to the pore air or pore water, and at what level between the two the average pressure lies. The consolidation stage of an effective stress triaxial test is carried out for two reasons. First, three specimens are tested and consolidated at three different effective pressures, in order to give specimens of different strengths which will produce widely spaced effective stress Mohr circles. Secondly, the results of consolidation are used to determine the minimum time to failure in the shear stage. The effective consolidation pressures (i.e. cell pressure minus back pressure) will normally be increased by a factor of two between each specimen, with the middle pressure approximating to the vertical effective stress in the ground. Effective stress triaxial tests are far less affected by sample size effects than undrained triaxial tests, but the problems of sampling in stoney soils still make multistage testing an attractive proposition under certain circumstances. The effectiveness of this technique in consolidated undrained triaxial testing has been reported by Kenney and Watson (1961), Parry (1968) and Parry and Nadarajah (1973). The consolidated drained triaxial compression test, with volume change measurement during shear is carried out in a similar sequence to the consolidated undrained test, but during shear the back pressure remains connected to the specimen which is loaded sufficiently slowly to avoid the development of excess pore pressures. The coefficient of consolidation of the soil is derived in the manner described above from the volume change measurements made during the consolidation stage. Thus the shear stage of a drained triaxial test can be expected to take between 7 and 15 times longer than that of an undrained test with pore pressure measurement. 100mm dia. specimens of clay may require to be sheared for as much as one month. Once shearing is complete, the results are presented as graphs of principal stress difference and volume change as a function of strain, and the failure Mohr circles are plotted to give the drained failure envelope defined by the parameters cd' and φd' The effective strength parameters defined by drained triaxial testing should not be expected to be precisely the same as those for an undrained test, since volume changes occurring at failure involve work being done by or against the cell pressure (Skempton and Bishop 1954). In practice the resulting angles of friction for cohesive soils are normally within 1—2°, and the cohesion intercepts are within 5 kN/m2. The results of tests on sands can vary very greatly (for example, Skinner 1969). Stiffness tests From the 1950s through to the early 1980s there has been a preoccupation in commercial soil testing with the measurement of strength with less emphasis being paid to the measurement of detailed stress—strain properties such as stiffness. This is reflected in both the 1975 and the 1990 editions of BS 1377, both of which fail to consider the measurement of stiffness.


March 2009 4-15

In most soils any discontinuities such as fissures will generally have a stiffness that is similar to that of the intact soil such that the intact soil stiffness may be used to predict with reasonable accuracy ground deformations and stress distributions. This means that laboratory triaxial tests on good quality ‘undisturbed’ specimens may yield adequate stiffness parameters for design purposes. However, conventional measurements of axial deformation of triaxial specimens, made outside the triaxial cell, introduce significant errors in the computation of strains.


4-16 March 2009

REFERENCES [1] American Association of State Highway and Transportation Officials (AASHTO). (1995). Standard specifications for transportation materials and methods of sampling and testing: part II: tests, Sixteenth Edition, Washington, D.C. [2] American Society for Testing & Materials. (2000). ASTM Book of Standards, Vol. 4, Section 08 and 09, Construction Materials: Soils & Rocks, Philadelphia, PA. [3] Bishop, A. W., and Henkel, D. J. (1962). The Measurement of Soil Properties in the Triaxial Test, Second Edition, Edward Arnold Publishers, Ltd., London, U.K., 227 p. [4] Bishop, A. W., and Bjerrum, L. (1960). “The relevance of the triaxial test to the solution of stability problems.” Proceedings, Research Conference on Shear Strength of Cohesive Soils, Boulder/CO, ASCE, 437-501. [5] Bishop, A. W., Alpan, I., Blight, G.E., and Donald, I.B. (1960). “Factors controlling the strength of partially saturated cohesive soils.”, Proceedings, Research Conference on Shear Strength of Cohesive Soils, Boulder/CO, ASCE, 503-532. [6] Clarke, B.G. (1995). Pressuremeters in Geotechnical Design. International Thomson Publishing/UK, and BiTech Publishers, Vancouver. [7] Deere, D. U., and Miller, R. P. (1966). Engineering classification and index properties of intact rock, Tech. Report. No. AFWL-TR-65-116, USAF Weapons Lab., Kirtland Air Force Base, NM. [8] Gibson, R. E. (1953). "Experimental determination of the true cohesion and true angle of internal friction in clays." Proceedings, 3rd International Conference on Soil Mechanics and Foundation Engineering, Zurich, Switzerland, 126-130. [9] International Society for Rock Mechanics Commission (1979). “Suggested Methods for Determining Water Content, Porosity, Density, Absorption and Related Properties.” International Journal Rock Mechanics. Mining Sci. and Geomechanics Abstr., Vol. 16, Great Britian, 141-156. [10] Jamiolkowski, M., Ladd, C. C., Germaine, J. T., and Lancellotta, R. (1985). “New developments in field and laboratory testing of soils.” Proceedings, 11th International Conference on Soil Mechanics & Foundation Engineering, Vol. 1, San Francisco, 57-153. [11] Littlechild, B.D., Hill, S.J., Statham, I., Plumbridge, G.D. and Lee, S.C. (2000). “Determination of rock mass modulus for foundation design”, Innovations & Applications in Geotechnical Site Characterization (GSP 97), ASCE, Reston, Virginia, 213-228. [12] LoPresti, D.C.F., Pallara, O., Lancellotta, R., Armandi, M., and Maniscalco, R. (1993). “Monotonic and cyclic loading behavior of two sands at small strains”. ASTM Geotechnical Testing Journal, Vol. 16 (4), 409-424. [13] LoPresti, D.C.F., Pallara, O., and Puci, I. (1995). “A modified commercial triaxial testing system for small strain measurements”. ASTM Geotechnical Testing Journal, Vol. 18 (1), 15-31. [14] Poulos, S.J. (1988). “Compaction control and the index unit weight”. ASTM Geotechnical Testing Journal, Vol. 11, No. 2, 100-108.


March 2009 4-17

[15] Richart, F. E. Jr. (1977). "Dynamic stress-strain relations for soils - State of the art report." Proceedings, 9th International Conference on Soil Mechanics and Foundation Engineering, Tokyo, 605-612. [16] Tatsuoka, F. and Shibuya, S. (1992). “ Deformation characteristics of soils & rocks from field & lab tests.” Report of the Institute of Industrial Science 37 (1), Serial No. 235, University of Tokyo, 136 p. [17] Tatsuoka, F., Jardine, R.J., LoPresti, D.C.F., DiBenedetto, H., and Kodaka, T. (1997). “Theme Lecture: Characterizing the pre-failure deformation properties of geomaterials”. Proceeedings, 14th International Conf. on Soil Mechanics & Foundation Engineering, Vol. 4, Hamburg, 2129-2164. [18] Tavenas, F., LeBlond, P., Jean, P., and Leroueil, S. (1983). “The permeability of natural soft clays: Parts I and II”, Canadian Geotechnical Journal, Vol. 20 (4), 629-660. [19] U.S. Department of the Interior, Bureau of Reclamation. (1973). Design of small dams, United States Government Printing Office, Washington, D.C. [20] U.S. Department of the Interior, Bureau of Reclamation (1960). Earth manual, United States Government Printing Office, Washington, D.C. [21] Woods, R. D. (1978). "Measurement of soil properties - state of the art report." Proceedings, Earthquake Engineering and Soil Dynamics, Vol. I, ASCE, Pasadena, CA, 91-178. [22] Woods, R.D. (1994). "Laboratory measurement of dynamic soil properties". Dynamic Geotechnical Testing II (STP 1213), ASTM, West Conshohocken, PA, 165-190. [23] Wroth, C. P., and Wood, D. M. (1978). "The correlation of index properties with some basic engineering properties of soils." Canadian Geotechnical Journal, Vol. 15 (2), 137-145. [24] Wroth, C. P. (1984). "The interpretation of in-situ soil tests." 24th Rankine Lecture, Géotechnique, Vol. 34 (4), 449-489. [25] oud, T.L. (1973). “Factors controlling maximum and minimum densities of sands”. Evaluation of Relative Density, STP 523, ASTM, West Conshohocken/PA, 98-112.


4-18 March 2009


CHAPTER 5 INTERPRETATION OF SOIL PROPERTIES


March 2009 5-i

Table of Contents


List of Table ........................................................................................................................... 5-ii


5.1 INTRODUCTION .......................................................................................................... 5-1

5.1.1 Reporting of Test Results ............................................................................... 5-1

5.2 COMPOSITION AND CLASSIFICATION ........................................................................... 5-2

5.2.1 Soil Classification and Geo-Stratigraphy ........................................................... 5-2

5.2.2 Soil Classification by Soil Sampling and Drilling ................................................ 5-2

5.2.3 Soil Classification by Cone Penetration Testing ................................................. 5-3

5.3 DENSITY ..................................................................................................................... 5-5

5.3.1 Unit Weight .................................................................................................. 5-5

5.3.2 Relative Density Correlations .......................................................................... 5-7

5.4 STRENGTH AND STRESS HISTORY ............................................................................... 5-11

5.4.1 Drained Friction Angle of Sands ..................................................................... 5-11

7.4.2 Pre-consolidation Stress of Clays ................................................................... 5-13

5.4.3 Undrained Strength of Clays and Silts ............................................................ 5-17

5.4.4 Lateral Stress State ...................................................................................... 5-20

5.5 FLOW PROPERTIES .................................................................................................... 5-21

REFERENCES ....................................................................................................................... 5-23


5-ii March 2009

List of Table


5.1 Representative Permeability Values for Soils 5-22

List of Figures


5.1 Delineation of Geostratigraphy and Soil & Rock Types by Drill & Sampling Methods 5-3

5.2 Factors Affecting Cone Penetrometer Test Measurements in Soils (Hegazy, 1998) 5-4

5.3 Chart for Soil Behavioral Classification by CPT (Robertson, Et Al., 1986) 5-5

5.4 Interrelationship between Saturated Unit Weight and In-Place Water Content Of Geo-Materials 5-6

5.5 Interrelationship between Minimum and Maximum Dry Densities of Quartz Sands 5-8

5.6 Maximum Dry Density Relationship with Sand Uniformity Coefficient 5-9

5.7 Relative Density Of Clean Sands From Standard Penetration Test Data 5-10

5.8 Relative Density Evaluations Of NC and OC Clean Quartz Sands from CPT Data 5-11

5.9 Typical Values of ø’ and Unit Weight for Cohesionless Soils 5-12

5.10 Peak Friction Angle Of Sands From SPT Resistance 5-12

5.11 Peak Friction Angle Of Un-Aged Clean Quartz Sands From Normalized CPT Tip Resistance 5-13

5.12 Representative Consolidation Test Results in Overconsolidated Clay 5-14

5.13 Trends for Compression and Swelling Indices in Terms of Plasticity Index 5-15

5.14 Ratio Of Measured Vane Strength To Preconsolidation Stress (Suv/P') Vs. Plasticity Index (Ip) (After Leroueil And Jamiolkowski. 1991) 5-15

5.15 Pre-consolidation Stress Relationship with Net Cone Tip Resistance from Electrical CPT 5-16

5.16 Relationship Between Pre-consolidation Stress and Excess Porewater Pressures from Piezocones 5-16

5.17 Relationship Between Pre-consolidation Stress and DMT Effective Contact Pressure in Clays 5-17

5.18 Relationship between Preconsolidation Stress and Shear Wave Velocity in Clays 5-17

5.19 Normalized Undrained Strengths for NC Clay under Different Loading Modes by Constitutive Model (Ohta, et al., 1985) 5-19

5.20 Undrained Strength Ratio Relationship with OCR and ' for Simple Shear Mode 5-20


March 2009 5-1

5 INTERPRETATION OF SOIL PROPERTIES

5.1 INTRODUCTION The results of the field and laboratory testing program must be compiled into a simplified representation of the subsurface conditions that includes the geo-stratigraphy and interpreted engineering parameters. Natural geo-materials are particularly difficult to quantify because they exhibit complex behavior and involve the actions and interactions of literally infinite numbers of particles that comprise the soil and/or rock mass. In contrast to the more "well-behaved" civil engineering materials, soils are affected by their initial stress state, direction of loading, composition, drainage conditions, and loading rate. Thus, the properties of soil and rock properties must be evaluated through a program of limited testing and sampling. In certain cases, the soil properties may be altered or changed using ground modification techniques. All interpretations of geotechnical data will involve a degree of uncertainty because of the differing origins, inherent variability, and innumerable complexities associated with natural materials. The interpretations of soil parameters and properties will rely on a combination of direct assessment by laboratory testing of recovered undisturbed samples and in-situ field data that are evaluated by theoretical, analytical, statistical, and empirical relationships. The application of empirical correlations and theoretical relationships should be done carefully, with due calibration and verification with the companion sets of laboratory tests, to ensure that proper site characterization is achieved. Notably, many interrelationships between engineering properties and field tests have developed separately from individual sources, with different underlying assumptions, reference basis, and specific intended backgrounds, often for a specific soil. 5.1.1 Reporting of Test Results Reporting of test results (field and laboratory) are presented in two basic forms. a. Factual Report b. Interpretative Report Factual Reports is a compilation of all the location plan of boreholes and test pits, borelogs, test pit logs, test results (field and laboratory) and photographs of site investigation activities without detailed interpretation of the test results. This report is basically presented by the S.I Contractor for their Client. Interpretative reports include the Factual Report as well as an interpretation of the test results by a geotechnical engineer/ expert to be used by the designers. This report can also be prepared by the S.I contractor by employing the services of a geotechnical engineer or it is prepared separately by the Client employing a geotechnical engineer depending on the nature of the site investigation contract. The interpretative report presents the interpretation of soil properties from in-situ tests and laboratory test for the analysis and design of foundations, embankments, slopes, and earth-retaining structures in soils. Correlation of properties to laboratory index tests and typical ranges of values are also provided to check the reasonableness of field and laboratory test results. Reference is made to relevant established documents and standards in order to familiarize with appropriate and more detailed directions on the procedures and methodologies, as well as examples of data processing and evaluation.


5-2 March 2009

5.2 COMPOSITION AND CLASSIFICATION Soil composition includes the relative size distributions of the grain particles, their constituent characteristics (mineralogy, angularity, shape), and porosity (density and void ratio). These can be readily determined by the traditional approach to soil investigation using a drilling and sampling program, followed by laboratory testing. The behavior of soil materials is controlled not only by their constituents, but also by less tangible and less quantifiable factors as age, cementation, fabric (packing arrangements, inherent structure), stress-state anisotropy, and sensitivity. In-situ tests provide an opportunity to observe the soil materials with all their relevant characteristics under controlled loading conditions. 5.2.1 Soil Classification and Geo-Stratigraphy In the field, there are three approaches to soil classification and the delineation of geo-stratigraphy, i.e., drilling and sampling, cone penetration, and flat plate dilatometer soundings. Testing by the cone and dilatometer, measure the in-situ response of soil while in its original position and environment, thus indicating a "soil behavioural" type of classification at the moment of testing. The field tests are primarily conducted by deployment of vertical soundings to determine the type, thickness, and variability of soil layers, depth of bedrock, level of groundwater and presence of lenses, seams, inclusions, and/or voids. 5.2.2 Soil Classification by Soil Sampling and Drilling Routine samplings involve the recovery of auger cuttings, drive samples, and pushed tubes from rotary-drilled boreholes. The boring may be created using solid flight augers (depth, z < 10 m), hollow-stem angers (z < 30 m), wash-boring techniques (z < 90 m), and wire-line techniques (applicable to 200 m or more). At select depths, split-barrel samples are obtained and a visual-manual examination of the recovered samples is sufficient for a general quantification of soil type. These 0.3-m long drive samples are collected only at regular 1.5-m intervals, however, and thus reflect only a portion of the subsurface stratigraphy. Less frequently, thin-walled undisturbed tube samples are obtained. More recently, sampling by a combination of direct-push and percussive forces has become available (e.g., geoprobe sampling; sonic drilling), whereby 25-mm diameter continuously-lined plastic tubes of soil are recovered. Although disturbed, the full stratigraphic profile can be examined for soil types, layers, seams, lenses, color changes, and other details.

March 200

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March 20

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5-6 March 2009

For the case of clays and silts above the water table, the soils may have degrees of saturation between 0 to 100%. Full saturation can occur due to capillarity effects and varies as the atmospheric weather. The identity relationship for total unit weight is:

γT = 1+wn

1+eo Gsγw

The estimation of unit weights for dry to partially saturated soils depends on the degree of saturation, as defined by (5.4) and (5.5).

Figure 5.4 Interrelationship between Saturated Unit Weight and In-Place Water Content Of Geo-Materials

The total overburden stress (σvo) is calculated from: σvo = ∑ T ∆z (5.6) which in turn is used to obtain the effective vertical overburden stress: σvo’ = σvo - uo (5.7) where the hydrostatic porewater pressure (uo) is determined from the water table.


March 2009 5-7

5.3.2 Relative Density Correlations The relative density (DR) is used to indicate the degree of packing of sand particles and applicable strictly to granular soils having less than l5 percent fines. The relative density is defined by:

DR = emax-eo

emax-emin (5.8)

where emax = void ratio at the loosest state and emin = void ratio at the densest state. The direct determination of DR by the above definition is not common in practice, however, because three separate parameters (eo, emax, and emin) must be evaluated. For a given soil, the maximum and minimum void states are apparently related (Poulos, 1988). A compiled database indicates (n = 304; r2 = 0.851; S.E. = 0.044): emin = 0.571 emax (5.9) For dry states (w = 0), the dry density is given as: d = Gs. γw/(l+e) and the relationship between the minimum and maximum densities is shown in Fig. 7.5 for a variety of sands. The mean trend is given by the regression line: d (min) = 0.808 d(max) (5.10) Laboratory studies by Youd (1973) showed that both emax and emin depend upon uniformity coefficient (UC = D60/D10), as well as particle angularity. For a number of sands (total n = 574), this seems to be borne out by the trend presented in Fig. 5.6 for the densest state corresponding to emin and d (max). The correlation for maximum dry density [ d (max)] in terms of UC for various sands is shown in Fig 5.7 and expressed by (n = 574; r2 = 0.730): d(max) = 9.8 [1.65 + 0.52 log (UC)] (5.11)


5-8 March 2009

Figure 5.5 Interrelationship between Minimum and Maximum Dry Densities of Quartz Sands. (Note: Conversion in terms of mass density and unit weight = 1 g/cc = 9.8 kN/m3 = 62.4 pcf)


March 2009 5-9

Figure 5.6 Maximum Dry Density Relationship with Sand Uniformity Coefficient (UC = D60/D10). (Note: Conversion In Terms Of Mass Density And Unit Weight: 1 G/Cc = 9.8 Kn/M3 = 62.4 Pcf)

From a more practical stance, in-situ penetration test data are used to evaluate the in-place relative density of sands. The original DR relationship for the SPT suggested by Terzaghi & Peck (1967) has been re-examined by Skempton (1986) and shown reasonable for many quartz sands. The evaluation of relative density (in percent) is given in terms of a normalized resistance [(N1)60], as shown in Fig. 5.7.

DR = 100 N1 60

60 (5.12)

where (N1)60 = N60/(σ.vo’)0.5 is the measured N-value corrected to an energy efficiency of 60%oand normalised to a stress level of one atmosphere. Note here that the effective overburden stress is given in atmospheres. In a more general fashion, the normalised SPT resistance can be defined by: (N1)60 = N60/(σvo’/pa)0.5 for any units of effective overburden stress, where pa is a reference stress = 1 bar ≈ 1 kg/cm2 ≈ 1 tsf ≈ 100 kPa. The range of normalized SPT values should be limited to (N1)60 < 60, since above this value, apparent grain crushing occurs due to high dynamic compressive forces. Additional effects of over-consolidation, particle size, and aging may also be considered, as these too affect the correlation (Skempton, 1986; Kulhawy & Mayne, 1990). A comparable approach for the CPT can be made based on calibration chamber test data on clean quartz sands (Fig. 5.8). The trends for relative density (in percent) of unaged uncemented sands are:


5-10 March 2009

Overconsolidated sands: DR = 100 qn

300 OCR0.2 (5.13a)

Normally-consolidated Sands: DR = 100 qn

300 (5.13b)

where qt1 = qc/(σvo’)0.5 is the normalized tip resistance with both the measured qc and the effective overburden stress are in atmospheric units. The relationship should be restricted to qt1 < 300 because of possible grain crushing effects. For any units of effective overburden stress and cone tip resistance, the normalized value is given by: qt1 = (qt/pa)/(σvo ‘/pa)0.5, where pa is a reference stress = l bar ≈ 1 kg/cm2 ≈ 1 tsf ≈ l00kPa.

Figure 5.7 Relative Density Of Clean Sands From Standard Penetration Test Data Note: Normalized Value (N1)60 = N60/(σ.Vo’)0.5 Where σVo’ is In Units Of Bars Or Tsf.


March 2009 5-11

Figure 5.8 Relative Density Evaluations Of NC and OC Clean Quartz Sands from CPT Data.

Note: Normalized resistance is qt1= qc/(σ’Vo)0.5 with stresses in atmospheres (1 Atm=1 Tsf=100 Kpa). 5.4 STRENGTH AND STRESS HISTORY The results of in-situ test measurements are convenient for evaluating the strength of soils and their relative variability across a project site. For sands, the drained strength corresponding to the effective stress friction angle (ø') is interpreted from the SPT, CPT, DMT, and PMT. For short-term loading of clays and silts, the undrained shear strength (cu) is appropriate and best determined from normalized relationships with the degree of over-consolidation. In this manner, in-situ test data in clays are used to evaluate the effective pre-consolidation stress (σp') from CPT, CPTu, DMT, and Vs, which in turn provide the corresponding over-consolidation ratios (OCR = σp'/σvo'). The long-term strength of intact clays and silts is represented by the effective stress strength parameters (ø’ and c’ = 0) that are best determined from either consolidated undrained triaxial tests with pore water pressure measurements, drained trail tests, or slow direct shear box tests in the lab. For fissured clay materials, the residual strength parameters (or’ and cry’ = 0) may be appropriate, particularly in slopes and excavations, and these values should be obtained from either laboratory ring shear tests or repeated direct shear box test series. 5.4.1 Drained Friction Angle of Sands The peak friction angle of sands (ø') depends on the mineralogy of the particles, level of effective confining stresses, and the packing arrangement (Bolton, 1986). Sands exhibit a nominal value of ø' due solely to mineralogical considerations that corresponds to the critical state (designated rocs'). The critical state represents an equilibrium condition for the particles at a given void ratio and effective confining stress level. For clean quartzite sands, a characteristic rocs' ≈ 33o, while a feldspathic sand may show øcs' ≈ 30o and a micaceous sandy soil exhibit øcs' ≈ 27o. Under many natural conditions, the sands are denser than their loosest states and dilatancy effects contribute to a peak ø' that is greater than øcs'. Fig. 5.9 shows typical values of ø' and corresponding unit weights over the full range of cohesionless soils.


5-12 March 2009

Figure 5.9 Typical Values of ø’ and Unit Weight for Cohesionless Soils. (NAVFAC DM 7.1, 1982)

The effective stress friction angle (ø') of sand is commonly evaluated from in-situ test data. The peak friction angles (ø') in terms of the (N1)60 resistances are presented in Fig. 5.10.

Figure 5.10 Peak Friction Angle Of Sands From SPT Resistance (Data From Hatanaka & Uchicla, 1996). Note: The Normalised Resistance Is (N1)60 = N60/(σVo’/Pa)0.5, Where Pa = 1 Bar = 1 Tsf = 100

Kpa


March 2009 5-13

The cone penetrometer can be considered a miniature pile foundation and the measured tip stress (qT) represented the actual end bearing resistance (qb). In bearing capacity calculations, the pile end bearing is obtained from limit plasticity theory that indicates: qb = Nq. σvo' where Nq is a bearing capacity factor for surcharge and depends upon the friction angle. Thus, one popular method of interpreting CPT results in sand is to invert the expression (Nq = qT/σvo') to obtain the value of φ' (e.g., Robertson & Campanella, 1983). One method for evaluating the peak φ’ of clean quartz sands from normalized CPT tip stresses is presented in Fig. 5.11.

Figure 5.11 Peak Friction Angle Of Un-Aged Clean Quartz Sands From Normalized CPT Tip Resistance. (Calibration Chamber Data Compiled By Robertson & Campanella, 1983).

7.4.2 Pre-consolidation Stress of Clays The effective preconsolidation stress σp', is an important parameter that governs the strength, stiffness, geostatic lateral stress state, and porewater pressure response of soils. It is best determined from one-dimensional oedometer tests (consolidation tests) on high-quality tube samples of the soil. Sampling disturbance, extrusion, and handling effects tend o reduce the magnitude of σp' from the actual in-place value. The normalised form is termed the overconsolidation ratio (OCR) and defined by: OCR = σp’/σvo’ (5.14) Soils are often over-consolidated to some degree because they are old in geologic time scales and have undergone many changes. Mechanisms causing over-consolidation include erosion, desiccation, groundwater fluctuations, aging, freeze-thaw cycles, wet-dry cycles, glaciation, and cementation.


5-14 March 2009

A representative e-log(σv’) curve obtained from one-dimensional consolidation testing on a marine clay is presented in Fig. 5.12. The observed pre-consolidation stress is seen to separate the recompression phase ("elastic strains") from the virgin compression portion (primarily "plastic strains") of the response. A check on the reasonableness of the obtained compression indices may be afforded via empirical relationships with the plasticity characteristics of the clay. A long-standing expression for the compression index (Cc) in terms of the liquid limit (LL) is given by (Terzaghi, et al., 1996): Cc = 0.009 (LL-10) (5.15) . In natural deposits, the measured Cc may be greater than that given by (5.15) because of inherent fabric, structure, and sensitivity. For example, in the case in Fig. 5.12 with LL = 41, (5.15) gives a calculated Cc = 0.33, vs. measured Cc = 0.38 in the oedometer.

Figure 5.12 Representative Consolidation Test Results in Overconsolidated Clay Statistical expressions for the virgin compression index (Cc) and the swelling index (Cs) from unload-reload cycles are given in Fig. 5.13 in relation to the plasticity index (PI). However, it should be noted that the PI is obtained on remoulded soil, while the consolidation indices are measurements on natural clays and silts. Thus, structured soils with moderate to high sensitivity and cementation will depart from these observed trends and signify that additional testing and care are warranted.

March 200

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5-16 March 2009

A direct correlation between the effective pre-consolidation stress and effective contact pressure (po-uo) measured by the flat dilatometer is given in Fig. 5.17, again noting that intact clays and fissured clays respond differently. The shear wave velocity (VS) can also provide estimates of σp', per Fig. 5.18. In all cases, profiles of σp' obtained by in-situ tests should be confirmed by discrete oedometer results.

Figure 5.15 Pre-consolidation Stress Relationship with Net Cone Tip Resistance from Electrical CPT

Figure 5.16 Relationship Between Pre-consolidation Stress and Excess Porewater Pressures from Piezocones


March 2009 5-17

Figure 5.17 Relationship Between Pre-consolidation Stress and DMT Effective Contact Pressure in Clays

Figure 5.18 Relationship between Preconsolidation Stress and Shear Wave Velocity in Clays. (Data from Mayne, Robertson, & Lunne, 1998)

5.4.3 Undrained Strength of Clays and Silts The undrained shear strength (su or cu) is not a unique property of soils, but a behavioral response to loading that depends upon applied stress direction, boundary conditions, strain rate, over-consolidation, degree of fissuring, and other factors. Therefore, it is often a difficult task to directly compare undrained strengths measured by a variety of different 1ab and field tests, unless proper

5-18

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March 20

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March 2009 5-19

Figure 5.19 Normalized Undrained Strengths for NC Clay under Different Loading Modes by Constitutive Model (Ohta, et al., 1985)

Based on extensive experimental data (Ladd, 1991) and critical state soil mechanics (Wroth, 1984), the ratio (su/σvo') increases with over-consolidation ratio (OCR) according to: (su/σvo’)OC = (su/σvo’)NC OCRA (5.16) where A ≈ 1- CS/CC and generally taken to be about 0.8 for unstructured and uncemented soils. Thus, if a particular shearing mode is required, it can be assessed using either Figs. 5.19 or 5.20 to obtain the NC value and equation (5-16) to determine the undrained strength for over-consolidated states. In many situations involving embankment stability analyses and bearing capacity calculations, the simple shear mode may be considered an average and representative value of the undrained strength characteristics, as shown by Fig. 5.21 and given by: (su/σvo’)DSS = ½ sin ’ OCRA (5.17)


5-20 March 2009

Figure 5.20 Undrained Strength Ratio Relationship with OCR and ' for Simple Shear Mode 5.4.4 Lateral Stress State The lateral geostatic state of stress (Ko) is one of the most elusive measurements in geotechnical engineering. It is often represented as the coefficient of horizontal stress Ko = σho'/σvo' where σho' = effective lateral stress and σvo' = effective vertical stress. A number of innovative devices have been devised to measure the in-place total horizontal stress (σho) including: total stress cell (push-in spade), self-boring pressuremeter, hydraulic fracturing apparatus, and the Iowa stepped blade. Recent research efforts attempt to use sets of directionalised shear wave measurements to decipher the in-situ Ko in soil formations. For practical use, it is common to relate the Ko state to the degree of overconsolidation, such as: K0 = (1 – sin ’) OCR sin ’ (5.18) which was developed on the basis of special laboratory tests including instrumented oedometer tests, triaxial cells, and split rings (Mayne & Kulhawy, 1982). Fig. 5.22 shows field data measurements of Ko for clays and sands.


March 2009 5-21

Fig. 5.22 Field Ko - OCR Relationships for (a) Natural Clays and (b) Natural Sand In general, the value of Ko has an upper bound value limited by the passive coefficient, Kp. The simple Rankine value is given by: Kp = tan2 (45° + ½ ’) = (1 + sin ’)/(1 + sin ’) (5.19) When the in-situ Ko reaches the passive value Kp, fissures and cracks can develop within the soil mass. This can be important in sloped masses since extensive fissuring is often associated with drained strengths that are at or near the residual strength parameters (φr' and cr' = 0). 5.5 FLOW PROPERTIES Soils exhibit flow properties that control hydraulic conductivity (k), rates of consolidation, construction behaviour, and drainage characteristics in the ground. Field measurements for soil permeability have been discussed previously and include pumping tests with measured drawdown, slug tests, and packer methods. Laboratory methods are presented in Section 4.6.5 and include falling head and constant head types in permeameters. An indirect assessment of permeability can be made from consolidation test data. Typical permeability values for a range of different soil types are provided in Table 5.1. Results of pressure dissipation readings from piezocone and flat dilatometer and holding tests during pressuremeter testing can be used to determine permeability and the coefficient of consolidation (Jamiolkowski, et al. 1985).

5-22

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March 20

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March 2009 5-23

REFERENCES [1] Barton, N.R. (1973). “Review of a new shear strength criterion for rock joints.” Engineering Geology, Elsevier, Vol. 7, 287-332. [2] Barton, N.R., Lien, R., and Lunde, J. (1974). "Engineering classification of rock masses for the design of tunnel support". Rock Mechanics, Vol. 6 (4), 189-239. [3] Barton, N.R. (1988). “Rock mass classification and tunnel reinforcement using the Q-system.”, Rock Classification Systems for Engineering Purposes, STP No. 984, ASTM, West Conshohocken, PA, 59-84. [4] Bieniawski, Z.T. (1984). Rock Mechanics Design in Mining and Tunneling. Balkema, Rotterdam, 272 p. [5] Bieniawski, Z. T. (1989). Engineering Rock Mass Classifications, John Wiley & Sons, Inc., New York. [6] Bieniawski, Z. T. (1972). “Propagation of brittle fracture in rock.” Proceedings., 10th U.S. Symposium. On Rock Mechanics., Johannesburg, South Africa. [7] Bishop, A. W., Alpan, I., Blight, G.E., and Donald, I.B. (1960). “Factors controlling the strength of partially saturated cohesive soils.”, Proceedings, Research Conference on Shear Strength of Cohesive Soils, Boulder/CO, ASCE, 503-532. [8] Bjerrum, L. (1972). “Embankments on soft ground.” Proceedings, Performance of Earth and Earth- Supported Structures, Vol. II, (Purdue Univ. Conf.), ASCE, Reston/VA, 1-54. [9] Bolton, M.D. (1986). "The strength and dilatancy of sands", Geotechnique, Vol. 36 (1), 65-78. [10] Bruce, D. A., Xanthakos, P. P., and Abramson, L. W. (1994). “Jet grouting”, Ground Control and Improvement, Chapter 8, 580-683. [11] Burland, J.B. (1989), "Small is beautiful: The stiffness of soils at small strains", Canadian Geotechnical Journal, Vol. 26 (4), 499-516. [12] Carter, M., and Bentley, S. P. (1991). Correlations of Soil Properties, Pentech Press Limited, London, U.K. [13] Casagrande, A., and Fadum, R. E. (1940). “Notes on soil testing for engineering purposes.” Publication 268, Graduate School of Engineering, Harvard University, Cambridge, Ma. [14] Cheney, R. S., and Chassie, R. G. (1993). “Soils and foundations workshop manual.” Circular FHWA HI-88-009, Federal Highway Administration, Washington D.C., 399. [15] Clarke, B.G. (1995). Pressuremeters in Geotechnical Design. International Thomson Publishing/UK, and BiTech Publishers, Vancouver. [16] Das, B. M. (1987). Advanced Soil Mechanics, McGraw-Hill Company, New York. [17] Das, B. M. (1990). Principles of Geotechnical Engineering,, PWS-Kent Publishing Company, Boston, MA, 665 p.


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[18] Deere, D. U., and Deere, D. W. (1989). Report Manual: Rock quality designation (RQD) after 20 years. [19] Duncan, J.M. and Chang, C.Y. (1970). “Nonlinear analysis of stress and strain in soils”. Journal of the Soil Mechanics & Foundation Division (ASCE) 96 (SM5), 1629-1653. [20] Federal Highway Administration (FHWA). (1985) “Checklist and guidelines for review of geotechnical reports and preliminary plans and specifications.” Report FHWA-ED-88-053, Washington D.C. [21] Federal Highway Administration (FHWA). (1989). “Rock slopes: design, excavation, stabilization.” Circular No. FHWA: TS-89-045, Washington, D.C. [22] Foster, R. S. (1975). Physical Geology, Merrill Publishing, Columbus, OH. [23] Franklin, J. A., and Dusseault, M. B. (1989). Rock Engineering, McGraw-Hill Company, New York. [24] Franklin, J. A. (1981). "A shale rating system and tentative applications to shale performance." Shales and Swelling Soils, Transportation Research Record 790, Transportation Research Board, Washington D.C. [25] gINT - gEotechnical INTegrator Software 3.2. (1991). “gINT, gEotechnical INTegrator Software 3.2, Documentation.” Geotechnical Computer Applications, Inc., Santa Rosa, California. [26] Goodman, R. E. (1989). Introduction to Rock Mechanics, Second Edition, John Wiley & Sons, Inc., New York, 562 p. [27] Hardin, B.O. and Drnevich, V.P. (1972). “Shear modulus and damping in soils”. Journal of the Soil Mechanics & Foundation Division (ASCE), Vol. 98 (SM7), 667-692. [28] Hassani, F.P., and Scoble, M.J. (1985). “Frictional mechanism and properties of rock discontinuities.” Proceedings, International Symposium on Fundamentals of Rock Joints, Björkliden, Sweden, 185-196. [29] Hilf, J. W. (1975). "Compacted fill." Foundation Engineering Handbook, H. F. Winterkorn and H. Y.Fang, eds., Van Nostrand Reinhold, New York, 244-311. [30] Hoek, E., and Bray, J. W. (1977). Rock Slope Engineering, Institution of Mining and Metallurgy, London, U.K. [31] Hoek, E., Kaiser, P.K., and Bawden, W.F. (1995). Support of Underground Excavations in Hard Rock, A.A. Balkema, Rotterdam, Netherlands. [32] Hoek, E. and Brown, E.T. (1998). “Practical estimates of rock mass strength”, International Journal of Rock Mechanics & Min. Sciences, Vol. 34 (8), 1165-1186. [33] Holtz, R. D., and Kovacs, W. D. (1981). An Introduction to Geotechnical Engineering, Prenctice-Hall, Inc., Englewood Cliffs, NJ. [34] Hough, B. K. (1969). Basic Soils Engineering, Ronald Press, New York.


March 2009 5-25

[35] Jaeger, J.C. and Cook, N.G.W. (1977). Fundamentals of Rock Mechanics, 2nd Edition, Science Paperbacks, Chapman & Hall, London, 585 p. [36] Jamiolkowski, M., Lancellotta, R., LoPresti, D.C.F., and Pallara, O. (1994). “Stiffness of Toyoura sand at small and intermediate strains”. Proceedings, 13th International Conference on Soil Mechanics & Geotechnical Engineering (1), New Delhi, 169-172. [37] Keaveny, J. and Mitchell, J.K. (1986). “Strength of fine-grained soils using the piezocone”. Use of In-Situ Tests in Geotechnical Engineering, GSP 6, ASCE, Reston/VA, 668-685. [38] Krebs, R. D., and Walker, E. D. (1971). "Highway materials." Publication 272, Department of Civil Engrg., Massachusetts Institute of Technology, McGraw-Hill Company, New York, 107. [39] Kulhawy, F.H. (1975). "Stress-deformation properties of rock and rock discontinuities", Engineering Geology, Vol. 9, 327-350. [40] Kulhawy, F.H. and Mayne, P.W. (1990). Manual on Estimating Soil Properties for Foundation Design. Report EPRI-EL 6800, Electric Power Research Institute, Palo Alto, 306 p. [41] KLadd, C.C., and Foott, R. (1974). "A new design procedure for stability of soft clay." Journal of Geotechnical Engineering, ASCE, Vol. 100 (3), 763-786. [42] Ladd, C.C. (1991). Stability evaluation during staged construction. ASCE Journal of Geotechnical Engineering 117 (4), 540-615. [43] Lambe, T.W. (1967). “The Stress Path Method.” Journal of the Soil Mechancis and Foundation Division, ASCE, Vol. 93 (6), Proc. Paper 5613, 309-331. [44] Lambe, T.W. and Marr, A.M. (1979). “Stress Path Method: Second Edition,” Journal of Geotechnical Engineering., ASCE, Vol. 105 (6), 727-738. [45] Lambe, T. W., and Whitman, R. V. (1979). Soil Mechanics: SI Version, John Wiley & Sons, Inc., New York, 553 p. [46] Lame, G. (1852). Lecons sur la theorie mathematique d'elasticite des corps solides, Bachelier, Paris, France (in French). [47] Littlechild, B.D., Hill, S.J., Statham, I., Plumbridge, G.D. and Lee, S.C. (2000). “Determination of rock mass modulus for foundation design”, Innovations & Applications in Geotechnical Site Characterization (GSP 97), ASCE, Reston, Virginia, 213-228. [48] Lupini, J.F., Skinner, A.E., and Vaughan, P.R. (1981). "The drained residual strength of cohesive soils", Geotechnique, Vol. 31 (2), 181-213. [49] Mayne, P.W. and Kulhawy, F.H. (1982). “K0-OCR relationships in soil”. Journal of Geotechnical Engineering, Vol. 108 (GT6), 851-872. [50] Mesri, G. and Abdel-Ghaffar, M.E.M. (1993). “Cohesion intercept in effective stress stability analysis”. Journal of Geotechnical Engineering 119 (8), 1229-1249. [51] Mitchell, J.K. (1993). Fundamentals of Soil Behavior, Second Edition, John Wiley & Sons, New York, 437 p.


5-26 March 2009

[52] NAVFAC, DM-7.1. (1982). "Soil Mechanics." Naval Facilities Engineering Command, Department of the Navy, Alexandria, VA. [53] Ng, C.W.W., Yau, T.L.Y., Li, J.H.M, and Tang, W.H. (2001). “Side resistance of large diameter bored piles socketed into decomposed rocks”, Journal of Geotechnical & Geoenvironmental Engineering Vol. 127 (8), 642-657. [54] Obert, L., and Duvall, W. I. (1967). Rock Mechanics and the Design of Structures in Rock, John Wiley & Sons, Inc., New York. [55] Ohta, H., Nishihara, A., and Morita, Y. (1985). “Undrained stability of Ko-consolidated clays.” Proceedings, 11th International Conference on Soil Mechanics & Foundation Engineering, Vol. 2, San Francisco, 613-616. [56] Patton, F. D. (1966). "Multiple modes of shear failure in rock." Proc., 1st International Congress on Rock Mechanics, Lisbon, Portugal, 509-13. [57] Peck, R. B., Hansen, W. E., and Thornburn, T. H. (1974). Foundation Engineering, John Wiley & Sons, Inc., New York, 514 p. [58] Pough, F.H. (1988). Rocks & Minerals. The Peterson Field Guide Series, Houghton Mifflin Company, Boston, 317 p. [59] Puzrin, A.M. and Burland, J.B. (1996). “A logarithmic stress-strain function for rocks and soils.” Geotechnique, Vol. 46 (1), 157-164. [60] Serafim, J. L. and Pereira, J. P. (1983). “Considerations of the geomechanics classification of Bieniawski.” Proceedings, International Symposium on Engineering Geology and Underground Construction, Lisbon, 1133-44. [61] Sheorey, P.R. (1997). Empirical Rock Failure Criteria. A.A. Balkema, Rotterdam, 176 p. [62] Singh, B. and Goel, R.K. (1999). Rock Mass Classification: A practical approach in civil engineering. Elsevier Science Ltd., Oxford, U.K., 267 p. [63] Skempton, A. W. (1957). Discussion on “The planning and design of new Hong Kong airport.” Proceedings, Institution of Civil Engineers, Vol. 7 (3), London, 305-307. [64] Soil Conservation Service (SCS). (1983). National soils handbook, Information Division, Washington, D.C. [65] Sowers, G.F. (1979). Introductory Soil Mechanics and Foundations, Geotechnical Engineering, Fourth Edition, Macmillan, New York. [66] Stagg, K. G., and Zienkiewicz, O.C. (1968). Rock Mechanics in Engineering Practice, John Wiley & Sons, Inc., New York. [67] Taylor, D. W. (1948). Fundamentals of Soil Mechanics, John Wiley & Sons, Inc., New York. [68] Terzaghi, K., and Peck, R. B. (1967). Soil Mechanics in Engineering Practice, John Wiley & Sons, Inc., New York, 729 p. [69] Terzaghi, K., Peck, R.B., and Mesri, G. (1996). Soil Mechanics in Engineering Practice, Second Edition, Wiley and Sons, Inc., New York, 549 p.


March 2009 5-27

[70] U.S. Department of the Interior, Bureau of Reclamation. (1973). Design of small dams, United States Government Printing Office, Washington, D.C. [71] U.S. Department of the Interior, Bureau of Reclamation (1960). Earth manual, United States Government Printing Office, Washington, D.C. [72] U.S. Department of the Interior, Bureau of Reclamation. (1986). "Soil classification handbook on Unified soil classification system." Training Manual No. 6; January, Geotechnical Branch, Washington, D.C. [73] Van Schalkwyk, A., Dooge, N., and Pitsiou, S. (1995). “Rock mass characterization for evaluation of erodibility”. Proceedings, 11th European Conference on Soil Mechanics and Foundation Engineering, Vol. 3, Copenhagen, Danish Geotechnical Society Bulletin 11, 281-287. [74] Vucetic, M. and Dobry, R. (1991). “Effect of soil plasticity on cyclic response”. Journal of Geotechnical Engineering, Vol. 117 (1), 89-107. [75] Way, D.S. (1973). Terrain Analysis, Dowden, Hutchingson & Ross, Inc., Stroudsburg, Pa. [76] Williamson, D.A. (1984). "Unified rock classification system." Bulletin of the Association of Engineering Geologists, Vol. XXI (3), 345-354 [77] Witczak, M.W. (1972). "Relationships between physiographic units and highway design factors." National Cooperative Highway Research Program: Report 132, Washington D.C. [78] Wittke, W. (1990). Rock Mechanics: Theory and Applications with Case Histories, Springer-Verlag, New York. [79] Wyllie, D. C. (1992). Foundations on Rock. First Edition, E&F Spon Publishers, Chapman and Hall, London, 333 p.


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DID MANUAL Volume 6

March 2009 i

Acknowledgements


DID MANUAL Volume 6

ii March 2009


Amend No

Page No


No Page No

Date of Admendment

DID MANUAL Volume 6

March 2009 iii

Table of Contents




Chapter 1 GEOMATICS AND LAND SURVEY SERVICES

Chapter 2 MAP PROJECTION

Chapter 3 TYPES OF SURVEY

Chapter 4 REFERENCES ON GEOMATICS AND LAND SURVEY SERVICES

Chapter 5 GEOGRAPHICAL INFORMATION SYSTEM (GIS)

Chapter 6 CHECKLIST FOR TERRAIN FEATURES

DID MANUAL Volume 6

iv March 2009


PART 3: ENGINEERING SURVEY

CHAPTER 1 GEOMATICS AND LAND SURVEY SERVICES


March 2009 1-i

Table of Contents



1.1 OVERVIEW ............................................................................................................... 1-1

1.2 FIELD OF GEOMATICS AND ENGINEERING SURVEYING ............................................... 1-1

1.3 APPLICATION AREAS ................................................................................................. 1-1

1.4 SOURCE OF MATERIAL FOR GEOMATIC PLANNING ...................................................... 1-1

1.5 PRINCIPLES OF SURVEYING EXERCISED BY SURVEYORS ............................................. 1-2

1.5.1 Basic Principles Adopted by Surveyors ........................................................... 1-2

1.5.2 Control ........................................................................................................ 1-2

1.5.3 Revision ................................................................................................................... 1-3

1.5.4 Economy and Accuracy ................................................................................. 1-4

1.5.5 The Independent Check ................................................................................ 1-4

1.5.6 Safeguarding ............................................................................................... 1-4

1.6 REFERENCES ............................................................................................................ 1-5


1-ii March 2009

List of Figures


1.1 Types of Traverse 1-3


March 2009 1-1

1 GEOMATICS AND LAND SURVEY SERVICES

1.1 OVERVIEW Planning and proposals for, and later, implementation of Department of Irrigation and Drainage project of various types have to take into consideration survey information provided by geomatics and land survey services. Geomatics is a fairly new term. It includes the tools and techniques used in land surveying for engineering works, remote sensing, Geographic Information System (GIS), Global Positioning System (GPS) and related forms of earth mapping. Originally, used in Canada, the term geomatics has been adopted by the International Organization for Standardization, the Royal Institution of Chartered Surveyors, the Institution of Surveyors Malaysia and many other international authorities. Some, especially the United States, prefer to use the term geospatial technology. The rapid progress and increased utilization of geomatics has been made possible by advances in computer technology, computer science and software engineering as well as advances in remote sensing technologies which provide imagery using space borne and air borne sensors. 1.2 FIELD OF GEOMATICS AND ENGINEERING SURVEYING a. Geodesy b. Surveying c. Mapping d. Positioning of structures e. Geomatic Engineering f. Navigation g. Remote Sensing h. Photogrammetry i. Geographic Information System j. Global Positioning System k. Geospatial Technology 1.3 APPLICATION AREAS a. The environment b. Land management c. Urban planning d. Subdivision planning in land development and land acquisition e. Infrastructure management f. Natural and infrastructure resource monitoring g. Coastal erosion management and mapping h. Natural disaster information for disaster risk reduction and response 1.4 SOURCE OF MATERIAL FOR GEOMATIC PLANNING a. In Malaysia the initial source for obtaining material and information to plan and then formulate

the term of reference and scope of work for proposals can be obtained from:- b. Topographic maps and aerial photographs from the Mapping Division of the Department of

Survey and Mapping [1] Department of Survey and Mapping Website: http://www.jupem.gov.my c. Cadastral Certified Plans and Cadastral Standard Sheets from the Cadastral Survey Division of

the Department of Survey and Mapping d. Thematic or geological maps from the Mineral and Geosciences Department Natural Resources

and Environment Ministry


1-2 March 2009

e. Malaysian Centre for Geospatial Data Infrastructure (MaCGDI) Ministry of Natural Resources and

Environment [2] MaCGDI Website:http:// www.mygeoportal.gov.my f. DigitalGlobe the provider of high resolution QuickBird Imagery. QuickBird’s high resolution

satellite imagery is available with resolution of 1.6 ft or 50cm panchromatic to 2ft or 70cm panchromatic, natural colors, colors infrared or 4-band pan sharpened [3] Digital Globe Website: http:// www.digitalGlobe.com. Digital Globe images has to be obtained through the Malaysian Centre for Remote Sensing (MACRES)

g. Combination in the supply of a mosaic assembled from Quick Bird Satellite Images supplied by Digital Globe and color aerial photographs supplied by the Department of Survey and Mapping Overlaid with Department of Survey and Mapping cadastral standard sheet information can be customized. e.g. Bertam area Kepala Batas

h. US Army Corps of Engineer Hydrographic Manual EM1110-2-1003 from the Web. (Chapter 17 – River Engineering and Channel Stabilization Surveys). [4] US Army Corps of Engineers website available by keying in “us army corps of engineers hydrographic survey manual” then click “EM1110-2-1003”

1.5 PRINCIPLES OF SURVEYING EXERCISED BY SURVEYORS 1.5.1 Basic Principles Adopted by Surveyors Users are informed that regardless of changes in techniques and equipment, the basic principles of surveying, which have been tested and proved over the years by geomatics and land surveyors remain and are applicable to all types of surveying. They are:-

a. Control comprising planimetric (Horizontal) and Height (Vertical) b. Revision c. Economy of Accuracy d. The independent check e. Save guarding

1.5.2 Control Any survey, whether large or small, depends upon the establishment of a carefully measured control framework which contains measured points linked with lines which encompass the whole area to be surveyed. The measured lengths and bearings of these straight lines, known as traverses, linking these series of points to form the various types of traverses are shown in Fig 1.1 below. Subsequent work is then fitted inside this framework and is adjusted to it. All TBMs should be connected by a closed leveling net which contain height points linked by survey lines which is tied to a minimum of 2 Survey Department Bench Marks (BM). Surveyors also check Azimuths or bearings reckoned from true north by solar observation of the sun at suitable intervals with maximum closing error of 1:4,000 for traverses within the horizontal control network (as a guide only). An open traverse is not acceptable unless it is double checked, both by angles and distances.


March 2009 1-3

Figure 1.1 Types of Traverse

1.5.3 Revision Whenever a survey is initiated, the methods and scope of works employed by the surveyor should be formulated in the light of the following requirements:

a. The requirements of the team of professionals who will be designing and subsequently implementing the project for the Department of Irrigation and Drainage. Checks should also be made that the requirement of another Department is taken into consideration e.g. the Ministry of Agriculture, Public Works Department or the Land Office resettlement plan.

b. It is important that a survey work done for one purpose may at some future date be used for a different purpose. The department concerned should anticipate this and consider whether, by some minor adjustment, the scope of works can be made more generally useful than the immediate needs.

c. It is important that all leveling or height control and connection work which include the establishment of hydrological stations are tied to Survey and Mapping Department Bench Marks (BM) and that Temporary Bench Marks (TBM) are established on permanent features at strategic locations within the proposed scheme for future use.

d. The field surveyor’s first task is to establish the horizontal and vertical control frameworks which are tied to the Survey Department Horizontal Datum for position and to the Land Survey Vertical Datum or the Chart Datum at the respective tide gauge stations for levels or TBM. Fitted within this framework are the supplementary control such as the DID proposed baseline, check line or secondary gridline where appropriate to pick up details of features and points contained in the Term of Reference (TOR)

KNOWN STATIONS

A. CLOSED LOOP TRAVERSE

KNOWN STATIONS KNOWN

STATIONS

B. CLOSED CONNECTION TRAVERSE

KNOWN STATIONS

C. OPEN TRAVERSE


1-4 March 2009

1.5.4 Economy and Accuracy It is important, before any field survey operation is started, to weigh the accuracy against the time, resources and costs. The greater the accuracy required, the greater the cost of operation. Since accuracy depends upon the elimination or reduction of errors, it is essential that the surveyor understands the nature of the errors and plans his works in such a way to reduce them to acceptable levels to meet the misclosure tolerances adopted. 1.5.5 The Independent Check In every survey operation it is the responsibility of a surveyor to do a check. It is best to employ a system which is completely self checking. Where this is not possible the check applied should be as independent as possible and not just a repetition of the previous operation. For example, if the measurement of the length is carried out, the check applied should be made by measuring the distance again using different unit of length or measuring in the reverse direction. In many cases a rough check is very useful and sometime all that is required. Computations which are not self checking should be completed by another survey staff including, using, if possible, methods other than those used. 1.5.6 Safeguarding Marks established by the field surveyor for the horizontal and vertical control framework should be as permanent as possible or easily re-established from nearby marks. Liaison with Agricultural Department may be considered during planning for topographical surveys to coordinate simultaneous concurrent activities to collect water and soil test samples to determine their suitability for crop cultivation. Hydrological stations for systematic collection of data such as rainfall, stream flow, maximum flood levels, tidal range, etc. should also be considered.


March 2009 1-5

REFERENCES [1] Department of Survey and Mapping website http://www.jupem.gov.my [2] Malaysian Centre for Geospatial Data Infrastructure (MaGDI) website http://www.mygeoportal.gov.my [3] Digital Globe for Satellite Imagery at website http://www.digitalglobe.com


1-6 March 2009

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CHAPTER 2 MAP PROJECTION


March 2009 2-i

Table of Contents



2.1 INTRODUCTION .......................................................................................................... 2-1

2.2 Map Projection Malaysia .................................................................................. 2-2

2.2.1 Rectified Skew Orthomorphic (RSO) Projection ..................................... 2-2

2.2.2 Cassini Soldner Projection ................................................................................ 2-2

2.2.3 WGS (World Geodetic System) 84 Ellipsoid ....................................................... 2-3

2.2.4 GDM 2000 or Geocentric Datum Malaysia 2000 .................................................. 2-3

2.3 REFERENCES ............................................................................................................... 2-5


2-ii March 2009

List of Figures


2.1 The Ellipsoid 2-1

2.2 RSO Grid Projection on Topographic Map 2-2

2.3 Peninsular Malaysia GPS Network 2-4


March 2009 2-1

2 MAP PROJECTION

2.1 INTRODUCTION • A map projection is used to portray all or part of the round Earth by transforming/projecting it

from a round surface (ellipsoid/spheroid) on to a plane or flat surface with some distortion. • Every projection has its own set of advantages and disadvantages. There is no “best”

projection. • The mapmaker must select the one best suited to the needs, reducing distortion of the most

important features. • Every flat map misrepresents the surface of the Earth in some way. No map can rival a globe in

truly representing the surface of the entire Earth. However, a map or parts of a map constructed from map projections can show one or more but never all of the following. True directions or bearings. True distances or scale. True areas. True shapes. Hence mapmaking is an art and science of trade-offs.

• Mapmakers and mathematicians have devised almost limitless equations to show the geographic image of the globe on paper. The mathematical model which is an approximation of the actual shape of the earth is commonly referred to as a spheroid or ellipsoid.

Elements of an ellipse

a = Semi Major Axis

b = Semi Minor Axis

f = Flattening = (a-b)/a

PP’ = Axis of revolution of the earth's ellipsoid

Figure 2.1 The Ellipsoid

• As shown in the Figure 2.1 above the surface of the earth is not a sphere but an irregular

changing shape, due to terrain features such as hills, mountains, valleys, rivers and the seas. This irregular surface has been approximated mathematically to that of an ELLIPSOID. Locations of topographic features on the curved surface of the ellipsoid earth are described in terms of latitude (Ø) Longitude (λ) and geodesic height (h). The ellipsoid parameters are expressed in terms of the semi major axis (a) and the flattening (f). These geographic coordinates which are then related mathematically to another system of mathematical coordinates on a flat/plane surface of a map are known as the rectangular Cartesian grid coordinates.

North pole P Geoid Equatorial Plane P1 Ellipsoid

ba


2-2 March 2009

2.2 MAP PROJECTION MALAYSIA 2.2.1 Rectified Skew Orthomorphic (RSO) Projection Rectified Skew Orthomorphic Projection has been adopted for the Topographic Maps produced by the Department of Survey and Mapping Malaysia. The result of this projection is the RSO Grid Coordinates. The Datum for this projection is KERTAU (Bukit Kertau Pahang). RSO projection was selected to suit the shape of Peninsular Malaysia. The limits of the projection are mainland Peninsular Malaysia and the close lying offshore islands. This RSO projection cannot be extended to include islands in the South China Sea, nor the East Malaysia states of Sabah and Sarawak. The East Malaysia states are covered by a second RSO projection. The Datum for this projection for the land below the wind is TIMBALAI (Timbalai Labuan). The mathematical theory on which the projection is based is found in the article “The Orthomorphic Projection of the Spheroid” by Brigadier M. Hotine CBE, published in the “Empire Survey Review” Vols VIII and IX Nos 62-65, particularly para 19 E.S.R. No. 64 of April 1947.

Figure 2.2 RSO Grid Projection on Topographic Map 2.2.2 Cassini Soldner Projection This projection was used extensively in Great Britain in the 19th Century where mapping was done by the respective counties (Majlis Perbandaran) whose areas are small. However it is not suitable for mapping of a nation as the projection is subjected to distortion of scales which increase progressively for areas whose distances increase from the central meridian of the ellipsoid. Similarly, the Cassini Soldner projection used in Peninsular Malaysia is on a state by state basis (except for the large state of Pahang which has 4 zones) by defining a central meridian and origin of projection for each of the states. Computation of cadastral coordinates for land title survey in Peninsular Malaysia based on the cassini soldner projection is very simple. It is based on the concept of selecting a fixed meridian and a point


March 2009 2-3

on the fixed meridian of the ellipsoid which acts as an origin. The coordinates of any point are then found as the length of perpendiculars from the point on the lot of a piece of land to the fixed meridian and the distance of the foot of the perpendiculars from the origin point. Geographical coordinates controlling cadastral surveys are computed on three separate datums namely the ASA datum (Bukit Asa) for the Southern Part of Peninsular Malaysia, the Kertau MRT datum for Terengganu, Perak and Kelantan and the Perak Datum (Gunong Hijau Larut) for Perlis, Kedah and Penang. However each state adopts its own coordinates system. 2.2.3 WGS (World Geodetic System) 84 Ellipsoid A unified global World Geodetic Reference System for relating the position of any feature or object on the surface of the earth become essential in the 1950s for several reasons:-

• International space science and the beginning of astronautics • The lack of inter-continental geodetic information • The inability of the large geodetic systems to provide a worldwide geographic coverage • Need for universal geographic reference system for global maps used for navigation, aviation

and geography or surveying

The new World Geodetic System called WGS 84 is currently the reference system used by the Global Positioning System. The WGS 84 originally used the GRS 30 reference ellipsoid but has undergone some minor refinements to meet high-precision calculations for the orbits of satellites. However these have little practical effect on typical topographic maps. Currently survey works by the Department of Survey and Mapping using GPS (Global Position System) is based on WGS 84 coordinates published by JUPEM (Jabatan Ukur dan Pemetaan) in 1994. 2.2.4 GDM 2000 or Geocentric Datum Malaysia 2000 The increasing usage of GPS by surveyors, engineers, navigators and other professionals especially those in GIS (Geographic Information System) applications, means that JUPEM has to provide geographically referenced map products which are compatible with worldwide usage of GPS without having to resort to lengthy computation steps which involves the transformation of coordinates such as follows:- (Ø λ h) < > (Ø λ h) < > (N, E.) < > (N, E) (WGS84) (MRT) (RSO) (Cassini) Future cadastral coordinate system will be based on the Geocentric Datum Malaysia 2000 or GDM2000. This system will replace the cassini soldner coordinates system mentioned above to facilitate the use of GPS. The GPS network which links all the GPS stations to form the Peninsular Malaysia Primary Geodetic Network for GDM2000 is depicted below.


2-4 March 2009

Figure 2.3 Peninsular Malaysia GPS Network

Longitude °E

Latitude °N


March 2009 2-5

REFERENCES [1] Department of Survey and Mapping website http://www.jupem.gov.my [2] United States Geological Survey website Map Projection Poster egsc.usgs.gov/isb/pubs/MapProjections/projections.html” [3] “The Orthomorphic Projection of the spheroid” Brigadier M. Hotine CBE in the Empire Survey Review vols VIII and IX Nos 62-65, particularly para 19 E.S.R. no. 64 of April 1947

CHAPTER 3 TYPES OF SURVEY


March 2009 3-i

Table of Contents


3 TYPES OF SURVEY ............................................................................................................... 3-1

3.1 INTRODUCTION .......................................................................................................... 3-1

3.2 CLASSIFICATION OF SURVEYS ..................................................................................... 3-1

3.2.1 Geodetic ...................................................................................................... 3-1

3.2.2 Plane .......................................................................................................... 3-1

3.2.3 Construction Surveys .................................................................................... 3-1

3.2.4 Topographic Mapping Surveys ....................................................................... 3-1

3.2.5 Basic Control (Geodetic) Surveys ................................................................... 3-2

3.2.6 Satellite Surveys .......................................................................................... 3-2

3.2.7 Hydrographic Surveys ................................................................................... 3-2

3.2.8 Land Surveys ............................................................................................... 3-2

3.2.9 Engineering Surveys ..................................................................................... 3-2

3.3 SURVEY NETWORKS .................................................................................................... 3-3

3.3.1 Basic Horizontal Control Network ................................................................... 3-3

3.3.2 Basic Vertical Control Network ....................................................................... 3-3

3.4 REAL TIME KINEMATIC (RTK) SURVEY .......................................................................... 3-3

3.5 LIDAR (Light Detection and Ranging) Airborne Mapping .................................................. 3-4

3.6 REFERENCES ............................................................................................................... 3-5

APPENDIX 3A-1 .................................................................................................................... 3A-1

APPENDIX 3A-2 .................................................................................................................... 3A-2


3-ii March 2009



March 2009 3-1

3 TYPES OF SURVEY

3.1 INTRODUCTION Surveying is the science of determining relative positions of points of geographical features on, under, or near the earth’s surface. These points may be cultural, hydrographic or terrain features on maps, or points needed to locate or layout roads, waterways, air fields or engineering structures of all kind 3.2 CLASSIFICATION OF SURVEYS Surveying which can be classed technically or functionally are described below:- 3.2.1 Geodetic A survey in which the figure and size of the mathematically created ellipsoidal shape of the earth is considered. It is applicable for large areas and long lines such as topographic mapping on a national scale. It is used for the precise location of higher order basic points in a control framework or net for controlling other lower order surveys. The Malaysia Primary Geodetic network and the GDM2000 Datum are described under “Basic Control (Geodetic) Surveys” item 3.2.5 and shown as Fig 2.3 Peninsular Malaysia GPS Network. 3.2.2 Plane In plane survey the curved surface of the earth is assumed to be flat. Currently cadastral survey for Issue Document of Title under the provision of the National Land Code Malaysia (Act 56 of 1965) is based on plane coordinates. For small areas, precise results may be obtained with plane-surveying methods, but the accuracy and precision of such results will decrease as the area surveyed is progressively increased in size. This is reflected in the need for each of the states in Peninsular Malaysia to have its own plane coordinate system except the very large state Pahang which has 4 zones. 3.2.3 Construction Surveys These surveys are conducted to obtain data essential to plan, design and estimate costs to locate or provide the layout points for implementing the construction of engineering structures. These surveys normally cover relatively small sites where the use of plane surveying techniques is adequate. 3.2.4 Topographic Mapping Surveys Topographic survey involves both air survey and field survey activities. Topographic surveys are conducted to establish horizontal and/or vertical positions of points which are then linked to similar distinctly identifiable points captured on aerial photograph for use by photogrammetric interpreters to compile topographic maps using computer aided mapping systems. Since the control stations are usually distributed over comparatively large areas their relative positions are determined by using point positioning by satellite techniques. Currently satellites from the GPS (Global Positioning System) which are being utilized globally are also widely used in Malaysia.


3-2 March 2009

3.2.5 Basic Control (Geodetic) Surveys Basic control survey provides positions, horizontal and or vertical, of geographic points on a terrain in a control framework to which supplementary surveys are adjusted. Most of these basic controls are limited to fit national mapping requirements and cannot be applied internationally. In Malaysia, these points are contained in two control network based on two local geodetic datum namely the Malayan Revised Triangulation (MRT) network for Peninsular Malaysia (West Malaysia) and the Borneo Triangulation 1968 (BT68) network for Sabah and Sarawak (East Malaysia). However, with the advent of new technologies such as the Global Positioning System (GPS) and Unified Geographic Information System (GIS) over large areas, the existing MRT and BT68 network have become outdated. A new Geocentric Datum of Malaysia (GDM2000) which fits into the global geodetic framework has been introduced to eventually replace the MRT and BT68. The GDM2000 datum contains the Peninsular Malaysia Primary Geodetic Network (PMPGN) of permanent GPS Stations established in 1998 for geodetic and scientific purposes. A similar East Malaysia Primary Geodetic Network (EMPGN) is being established. 3.2.6 Satellite Surveys Satellite surveys employ the use of artificial earth satellites as a means of extending geodetic control systems. These positioning of points on the ground in a geodetic control system are being conducted using artificial earth satellites in the Global Positioning System (GPS) for long line surveys where the distance between stations is a few hundred kilometers apart. They are used for conducting worldwide surveys for intercontinental, inter-datum and inter-island geodetic ties. Topographic and basic control surveys are frequently conducted with satellite surveys. Special project instructions are written to detail methods, techniques, equipment and procedures to be used in these surveys. 3.2.7 Hydrographic Surveys A survey made in relation to any considerable body of water, such as a strip of part of the sea along the coast, a bay, harbour, lake or river for the purpose of determination of channel depths for navigation, location of rocks, sand bars, and in the case of rivers for flood mitigation control, hydro-electric power generation, navigation of boats, water supply and water storage. 3.2.8 Land Surveys Land surveying embraces survey operations to locate and monument the boundaries of a property to meet the requirement of Land Laws relating to land and land tenure in the National Land Code (Act 56 of 1965). In the case where alienated land is acquired for construction works such as flood mitigation projects land survey has to be conducted to meet the requirement of the Land Acquisition Act. Land survey is commonly referred to as Cadastral Survey. 3.2.9 Engineering Surveys It is executed for the purpose of obtaining information which is essential for planning an engineering project or proposed development and estimating its cost. The survey information may, in part, be in the form of an engineering survey map.


March 2009 3-3

3.3 SURVEY NETWORKS Horizontal and vertical survey control within a country like Malaysia was established by a network of control arcs, which are all referenced to a single datum and are therefore linked in position and elevation to each other, regardless of their distance apart. These networks for topographic mapping are referenced to the KERTAU Datum for the Malayan Revised Triangulation (MRT) network in Peninsular Malaysia and the TIMBALAN Datum for the Borneo Triangulation 1968 (BT68) network in the Sabah and Sarawak states of East Malaysia. 3.3.1 Basic Horizontal Control Network The horizontal control for mapping was established by connecting a mixed series of stations (geodetic, primary, secondary and tertiary stations) by a combination of precise electronic distance measuring techniques (Geodimeter) and first order astronomical observation to form the Malaysian geodetic net covering Peninsular Malaysia. The stations in the network were then transformed into the RSO coordinates system. This network is being replaced by the GDM2000 network, shown in Fig 2.3, which has been established using GPS satellite point positioning techniques to fit it into a global geodetic framework. This network is termed Malaysia Primary Geodetic Network (PMPGN) and the East Malaysia Primary Geodetic Network (EMPGN). 3.3.2 Basic Vertical Control Network This control was established to provide orthometric (mean sea level) heights in the national height system in the configuration of leveling networks. The datum for orthometric leveling in Peninsular Malaysia is Bench mark No. B0169 Height 3.863 metres above Mean Sea Level (MSL) located at the back of the tide gauge station on Warf No. 25 North Port, Port Klang. Hydrographic survey for design of marine structures may require the heights to be tied to the Chart Datum used in Nautical Charts. In such situations the Orthometric (Mean Sea Level) heights relative to the Chart Datum available from the Hydrographic Division of the Royal Malaysia Navy has to be obtained. Fig 4.1 Survey Datum shows the relationship between the Chart Datum and Land Survey Datum. 3.4 REAL TIME KINEMATIC (RTK) SURVEY The Geodesy Section, Department of Survey and Mapping Malaysia provide Real Time Kinematic (RTK) Virtual Reference Station (VRS) technique which extends the use of RTK to the whole of Peninsular Malaysia by the establishment of a network containing GPS reference stations over the whole of Peninsular Malaysia. This service, which attracts a standard fee, is provided by the Malaysia Real-Time Kinematic GPS Network System (MyRTnet), for users to conduct dynamic GPS Survey to meet applications below:- • Geomatics • Deformation Monitoring • Scientific Research • Surveying • Construction • Navigation • Mapping and GIS (Geographic Information System) • Location Based Services RTK VRS networking exploits the concept of all users sharing a common GPS coordinate control framework and it significantly reduces systematic errors and extends the operating range with improved accuracy requiring less time. It is surveying where users do not have to set-up their own base stations


3-4 March 2009

Appendix 3A-1 shows in general the concept on functioning of the RTK network together with cellular phone (gsm) communication to obtain the geographical position of a map or engineering feature to an accuracy of +/- 2 to 3 cm. 3.5 LIDAR (Light Detection and Ranging) Airborne Mapping Light Detection and Ranging (LIDAR) is an airborne mapping technique which uses Laser to measure the distance between the aircraft and the terrain of the ground. Airborne LIDAR systems can broadly be classified into 3 main types: Wide Area Mapping using fixed wing aircrafts, Corridor Mapping Systems mounted on helicopters and bathymetric mapping systems using either one of these two airborne platforms. A typical airborne LIDAR system coupled with a Global Positioning System (GPS) and an Inertial Navigation System (INS) allow the user to capture geo-referenced “Points” of ground features to produce highly accurate Digital Elevation Models (DEMs) either day or night in a variety of weather conditions. The LIDAR system acquires data along a corridor that can be up to 600 metres wide. These very accurate elevation data have a variety of uses, such as the generation of contour lines, beach profiles and modeling terrain for 3D applications. Data acquired using LIDAR systems are often used in conjunction with data from other remote sensing instruments; including spectral and thermal imaging system and high resolution video and digital aerial cameras to produce digitally rectified images or orthophotographs. More information on LIDAR is contained in item 4.17 and Appendix 3A-2.


March 2009 3-5

REFERENCES [1] Department of Survey and Mapping website http://www.jupem.gov.my [2] “The Orthomorphic Projection of the spheroid” Brigadier M. Hotine CBE in the Empire Survey Review vols VIII and IX Nos 62-65, particularly para 19 E.S.R. no. 64 of April 1947 [3] GDM2000 Geodesy Section, Department of Survey and Mapping website http://geodesi.jupem.gov.my


3A-1 March 2009

APPENDIX 3A-1 Illustration on Point Positioning for using Satellite and RTK (Real Time Kinematic) Networking

JUPEM - Jabatan Ukur Dan Pemetaan Malaysia (Department of Survey and Mapping Malaysia) MyRTKnet - Malaysia Real Time Kinematic GPS network system control center RTCM - Radio Technical Commission for Maritime Services (RTCM) Standard for mobile

phone communication to enable the field surveyor to obtain the real time position of a point to an accuracy of +/- 2 to 3 cm from myRTKnet

JUPEM GPS reference Station - A GPS station within the JUPEM Network of RTK GPS reference stations


March 2009 3A-2

APPENDIX 3A-2

LIDAR (Light Detection and Ranging) Airborne System comprising • Laser Scanner • GPS (Global Positioning Satellite) Receiver • IMU (Inertial Measurement Unit)

IMU

Airborne LIDAR System


3A-3 March 2009


CHAPTER 4 REFERENCES ON GEOMATICS AND LAND SURVEY SERVICES


March 2009 4-i

Table of Contents


List of Figures ........................................................................................................................ 4-iii

4.1 INTRODUCTION .......................................................................................................... 4-1

4.2 POINT POSITIONING OF A FEATURE USING SATELLITE AND RTK ........................................ (JADUAL 2001 ITEM 1.4). ............................................................................................. 4-1

4.3 PLANIMETRIC (HORIZONTAL TRAVERSING) CONTROL AND CONNECTION .......................... (JADUAL 2001 ITEM 1.5). ............................................................................................. 4-1

4.4 HEIGHT (VERTICAL) CONTROL AND CONNECTION (JADUAL 2001 ITEM 1.6). .................. 4-2

4.5 LEVELING BENCH MARKS (BM) OR MONUMENTATION (JADUAL 2001 ITEM 1.7) .............. 4-3

4.6 TOPOGRAPHICAL SURVEY (JADUAL 2001 ITEM 2.10 AND ITEM 7.9) ................................ 4-4

4.7 GRID SURVEY (JADUAL 2001 ITEM 7.9.2 AND ITEM 2.2 IN KEMENTERIAN ........................... KEWANGAN KHAZANAH MALAYSIA LETTER REFERENCE (K&B)(8.09)735/3/1 JD.3(13) ........... DATED 13TH JANUARY 1984). ...................................................................................... 4-4

4.8 SETTING-OUT SURVEY (JADUAL 2001 ITEM 8.10 AND 8.13). .......................................... 4-4

4.9 SURVEY OF EXISTING WATERWAYS, CANALS AND DRAINS ................................................ (JADUAL 2001 ITEM 8.11 AND 3.10.2) .......................................................................... 4-4

4.10 STRIP SURVEY TO MAP DETAILS AND SPOT LEVELS (JADUAL 2001 ITEM 4.9 AND 8.9) .... 4-4

4.11 PREPARATION OF LAND ACQUISITION PLANS (JADUAL 2001 ITEM 8.14 & 1.11 ................... AND REGULATION 1991 ITEM 3(B). .............................................................................. 4-5

4.12 EFFECT OF ADVANCE OR RETREAT OF THE BED OF ANY RIVER OR SEA .......................... 4-5

4.13 TRANSFORMATION OF COORDINATES AND MAP PROJECTIONS IS NEEDED DUE ................. TO THE USE OF VARIOUS GEOGRAPHIC REFERENCE SYSTEMS (JADUAL 2001 ..................... ITEM 8.16 AND 1.13). .................................................................................................. 4-5

4.14 AIR SURVEY MAPPING TECHNIQUE FOR PRODUCING ENGINEERING SURVEY PLANS (JADUAL 2001 ITEM 11) .............................................................................................. 4-6

4.14.1 Limitation of Air Survey ............................................................................... 4-7

4.15 HYDROGRAPHIC SURVEY FOR TERRITORIAL WATERS AND INLAND WATER BODIES (JADUAL 2001 ITEM 14 PART V) ................................................................................................. 4-7

4.16 LOCATING OF CROSS-SECTION PROFILES FOR HYDRAULIC ENGINEERING (JADUAL 2001 ITEM 14.9 PART V) ...................................................................................................... 4-8

4.16.1 Mixed Survey Methods ................................................................................ 4-8

4.16.2 Guidance to Surveyors on Cross-Section Locations......................................... 4-8

4.16.3 Guidelines on Locating Cross-Sections .......................................................... 4-8

4.16.4 Additional Guidelines on Cross-Section Profiles .............................................. 4-9

4.16.5 Cross-Sections Adjacent to Bridges or Culverts (Jadual 2001 Item 3 Part I) ... 4-10

4.17 LIDAR (LIGHT DETECTION AND RANGING) AIRBORNE MAPPING .................................. 4-10

4.18 REFERENCES ............................................................................................................. 4-12

APPENDIX 4A-1 .................................................................................................................... A4-1


4-ii March 2009

APPENDIX 4A-2 .................................................................................................................. A4-45

APPENDIX 4A-3 .................................................................................................................. A4-57

APPENDIX 4A-4 .................................................................................................................. A4-61

APPENDIX 4A-5 .................................................................................................................. A4-64

APPENDIX 4A-6 .................................................................................................................. A4-69

APPENDIX 4A-7 .................................................................................................................. A4-72


March 2009 4-iii

List of Figures


4.1 Survey Datum 4-3

4.2 Typical Cross-Section Configuration 4-9

4.3 Cross-Section Locations at a Bridge or Culvert 4-10


4-iv March 2009



March 2009 4-1

4 REFERENCES ON GEOMATICS AND LAND SURVEY SERVICES

4.1 INTRODUCTION Guides to components of surveying which are important for ascertaining cost estimates and specifying scope of survey works in the planning of any Department of Irrigation and Drainage project are currently guided by contents in the following references. Updates which are issued from time to time should be applied where relevant in the future to these references.

a. Kelulusan Kadar Baru Pengiraan Kos Perkhidmatan Perunding Bidang Ukur Tanah bagi Projek-Projek Kerajaan. Perbendaharaan Kementerian Kewangan Malaysia letter reference S(K&B)(8.09)735/9-24 Sj.5.Jld.3 (11) dated 29th March 2005.

b. Jadual Fee Ukur Kejuruteraan 2001 (Pindaan Kepada Jadual Fee Ukur Kejuruteraan 1980). Please see Appendix 4A-1.

c. Peraturan-peraturan Jurukur Tanah Berlesen (Pindaan) 1997 (Kadar Bayaran Upah Ukur untuk Ukuran Hakmilik) Akta Jurukur Tanah Berlesan 1958 P.U. (A) 169. Please see Appendix 4A-2.

d. Surat Perkeliling Perbendaharaan Bil.8 Tahun 2006 on Peraturan Perolehan Perkhidmatan Perunding reference S/K.KEW/PK/1100/000000/10/31 Jld.21 (5) dated 6th November 2006. Please see Appendix 4A-3.

e. Chapter 17 River Engineering and Channel Stabilization Surveys EM1110-2-1003 US Army corps of Engineers Hydrographic Survey Manual.

f. BQ Example - Cost Estimate for Survey of Existing Route of Waterways, Canals and Drains. Please see Appendix 4A-4.

g. BQ Example - Cost Estimate for Hydrographic Survey of Territorial Waters and Inland Water Bodies. Please see Appendix 4A-5.

4.2 POINT POSITIONING OF A FEATURE USING SATELLITE AND RTK (JADUAL

2001 ITEM 1.4). The Global Positioning System (GPS) is currently the only fully functional Global Navigation Satellite System (GNSS). Utilizing a constellation of at least 24 medium Earth Orbit Satellites that transmit precise microwave radio signals, the system enables a GPS receiver to determine the Position of a point or location on or above the surface of the earth. The GPS radio receiver has become a widely used aid to navigation worldwide and a useful tool, among many others, map making and Land Surveying. GPS equipment used by surveyors incorporates techniques and augmentation methods to improve accuracy and error sources inherent to operation of GPS. Example of augmentation systems includes Differential GPS or RTK (Real-Time Kinematic) surveying illustrated at Appendix 3A-1. In Malaysia RTK survey service for a fee is provided by logging on to myRTKnet located at the Geodesy Section of the Department of Survey and Mapping. 4.3 PLANIMETRIC (HORIZONTAL TRAVERSING) CONTROL AND CONNECTION

(JADUAL 2001 ITEM 1.5). Planimetric cntrol and connection is a technique used for determining the relative horizontal positions (x, y coordinates) of cultural, hydrographic or terrain features for mapping or points needed to plan and subsequently locate positions or layout accurately bunds, canals, soil investigation boreholes, roads, waterways and drainage structures, of all kinds. It comprises a series of points on features surveyed. Hence it comprises:


4-2 March 2009

a. Connection to Survey Department Horizontal Datum which provides scale, position and

azimuth control for establishing boundary marks shown on land title survey or cadastral survey plans to meet issue document of titles to land based on the Cassini Soldner projection.

b. And then the proposed or existing routes or alignment, identified by the Department of Drainage and Irrigation. Issue to be considered here are land with various category of titles and ownership which have to be obtained from the Cadastral Division Department of Survey and Mapping, the Land Office and sometimes direct objection from the affected land owner himself.

4.4 HEIGHT (VERTICAL) CONTROL AND CONNECTION (JADUAL 2001 ITEM 1.6). Height controls and connection to determine the spot level of a feature includes:

a. Connection to Survey and Mapping Department Bench Marks (BM) based on the Land Survey Datum (LSD) and now known as the National Vertical Geodetic Datum (NGVD) which is located at a tide gauge station sited in Port Klang, Selangor

b. Connection to the CHART DATUM which is traditionally referred to as the Admirably Chart Datum. These datums are located at Tidal Stations, mainly jetties or ports along the coast. Appendix 3A-1 attached contains a list of the existing Tidal Stations.

c. Occasionally connection to both the LSD and the Admirably Chart Datum has to be related for marine navigation structures such as a fishing jetty or port. An example of this is depicted in the diagram below which shows the Chart Datum is 1.7m below the Land Survey Datum.

d. And then along the proposed or existing routes or alignment identified by the Drainage and Irrigation Department


March 2009 4-3

Figure 4.1 Survey Datum 4.5 LEVELING BENCH MARKS (BM) OR MONUMENTATION (JADUAL 2001 ITEM

1.7) A Bench Mark is a relatively permanent object natural or artificial, bearing a marked point normally a brass bolt set in concrete with a bench mark number inscribed. The elevation or the height of the point above or below the Land Survey Datum (LSD) or National Geodetic Vertical Datum (NGVD) has to be purchased from Geodesy Section, Department of Survey and Mapping. Establishment of subsidiary marks or monuments related to the Department of Survey and Mapping Bench Marks by conducting Height Control and Connection Surveys are known as:-

a. Temporary Bench Marks (TBM) • Plan of a TBM marker on Normal surface is shown in Appendix 4A-6. • Plan of a TBM marker on hard surface is shown in Appendix 4A-7.

b. Intersection Point Marks (IP) c. Reference Marks (RM)

NAU

TICA

L O

R A

DM

IRAL

TY CH

ART

DAT

UM

TIDAL REFERENCE FOR PULAU SIBU

31.847 m (BM S1150) Above L.S.D

00m (L.S.D)/(M.S.L)

-1.700m (Chart Datum)

32.547m (BM S1150) Above Chart Datum

1.700m

00 (Chart Datum)

Chart Datum is 1.700M below Land Survey Datum (L.S.D) at Survey Department Bench Mark (BM S1150)


4-4 March 2009

4.6 TOPOGRAPHICAL SURVEY (JADUAL 2001 ITEM 2.10 AND ITEM 7.9) Topographic Surveys often known as Engineering Surveys are conducted to establish horizontal (x, y) and/or vertical (h or z) positions of points of all natural and manmade features to produce a geographical details and contour map over a large area. Topographic maps supply a general image of the earth’s surface namely roads, rivers, buildings, often the nature of the vegetation, the contour together with spot levels and names of various surveyed objects. The main supplier of topographic map is the Department of Survey and Mapping Malaysia 4.7 GRID SURVEY (JADUAL 2001 ITEM 7.9.2 AND ITEM 2.2 IN KEMENTERIAN

KEWANGAN KHAZANAH MALAYSIA LETTER REFERENCE (K&B)(8.09)735/3/1 JD.3(13) DATED 13TH JANUARY 1984).

This survey is special to projects where the difference in spot levels is very important and critical. It is specified for survey of aircraft runway construction or other flat surface. This type of survey is not suitable for undulating or hilly area covered by overgrown vegetation. 4.8 SETTING-OUT SURVEY (JADUAL 2001 ITEM 8.10 AND 8.13). This survey, also known as construction setting out survey, is executed before construction works can start. The setting comprise x and y coordinates of the following:-

a. Centre line of proposed route from IP to IP (Intersection Points) b. Right of Way (ROW) of the waterway, canal or drain reserve based on the approved pre-

computation plan. c. Intersection Points (IP) (Jadual 2001 Item 8.10) d. Pegging of positions of Piling Points based on pre-computation plan from engineering layout

plan 4.9 SURVEY OF EXISTING WATERWAYS, CANALS AND DRAINS (JADUAL 2001

ITEM 8.11 AND 3.10.2) This survey covers the area within the banks or the designated or gazette reserve for the irrigation canal or waterway to show the alignment, longitudinal section and the cross-sections. It also includes the area within the specified Right of Way (ROW) shown on the approved pre-computation plan. When the reserve is not specified the outer limits of the alignment is within 50m from the banks of river or drain or canal. If the water depth of the waterways, drains and canal at the time of survey is more than 1 meter then Jadual 2001 item 8.11 specification (viii) and item 3.10.2 applies or alternatively Hydrographic Survey for Inland Water Bodies under Paragraphs 4.15 (Jadual 2001 Item 14 Part V) and 4.16 (Jadual 2001 Item 14.9.1 Part V) is applicable. If the width of the cross-sections or the intervals is more or less than 50 metres then the fees shall be increased or decreased proportionately (specification (vii) Jadual 2001 item 8.11) 4.10 STRIP SURVEY TO MAP DETAILS AND SPOT LEVELS (JADUAL 2001 ITEM 4.9

AND 8.9)

The strip comprises topographic details and spot levels survey of long narrow stretches of areas or corridors which are beyond the banks or overbanks and flood plains of a waterway or along the coast.


March 2009 4-5

4.11 PREPARATION OF LAND ACQUISITION PLANS (JADUAL 2001 ITEM 8.14 &

1.11 AND REGULATION 1991 ITEM 3(B). These are approved pre-computation plans based on the Right of Way (ROW) plans or any other area to be acquired for the implementation of the project for the proposed waterway, canal or drain. On being surveyed the ROW will eventually become the designated drainage reserve alignment to be maintained by the Drainage and Irrigation Department. Land acquisition is a socially sensitive, very costly, tedious and long drawn process which entails the following:-.

a. Preparation of Land Acquisition Plans which comprise:- • Purchasing certified plans (Pelan Akui) and Cadastral (Standard) Sheets from the

Department of Survey and Mapping for the compilation/preparation of Land Acquisition (LA) Plans.

• Search for Qualified Titles (Hakmilik Sementara) and Approved LA Plans at the Land Office or other Government Department.

b. Land Acquisition Plans normally compiled on the same scale as the Survey Department cadastral sheet shall show:- • Lot boundaries with bearings and distances within the surveyed corridor or strip

(proposed alignment/ROW) • Lot numbers of lots to be acquired • Lot areas with details on portion to be acquired and the left over balance • Status and category of land use and crops • Houses and other as-built features affected by the Acquisition

c. Finalized Land Acquisition Plans are updated from:-

• Revision/amendment of ROW by consulting engineer • Comments by the Department of Drainage and Irrigation • Up-to date information on change in status of land received from the Land Office • Objection from Land owner during field survey work to demarcate the ROW/alignment of

the future waterway reserve or alignment.

d. R.S. (Requisition for Survey) Plan. The approved Pre-computation plan for Land Acquisition which is attached to the Requisition for Survey (Permintaan Ukur) letter by the Land Office to the Department of Survey Mapping is known as the R. S. Plan.

4.12 EFFECT OF ADVANCE OR RETREAT OF THE BED OF ANY RIVER OR SEA Frequently while conducting survey for Land Acquisition we come across a situation where part a of privately owned land along river banks are lost through erosion by the action of flood water. Similarly land along the opposite bank, especially on bends, may also gain land through accretion by the action of flooding. Such lands, as per provision of Section 49 of the National Land Code (Act 56 of 1965), shall become State land. 4.13 TRANSFORMATION OF COORDINATES AND MAP PROJECTIONS IS NEEDED

DUE TO THE USE OF VARIOUS GEOGRAPHIC REFERENCE SYSTEMS (JADUAL 2001 ITEM 8.16 AND 1.13).

Coordinates in a common geographically referenced system is needed to provide information on the location of a position of a feature for navigation, point of Interest or geographic information system. An example on the request for transformation of coordinates is the experience with the Sungai Muda Flood Mitigation Project stretching from Jambatan Merdeka to Kuala Muda where different coordinates are being used.


4-6 March 2009

a. The various coordinates used are:-

• The plane cadastral coordinates (x and y coordinate) on the Kedah side of Sungai Muda is based on the Cassini Solder projection from the Central Meridian of the Everest (Modified) Ellipsoid at Gunung Perak

• The plane cadastral coordinates in Pulau Pinang side of Sungai Muda is based on the Cassini Solder projection from the Central Meridian of the Everest (Modified) Ellipsoid at Gun Hill.

• The coordinates required by the Civil and Structural Consultants is that based on the RSO (Rectified Skew Orthomorphic) projection. The RSO projection is used for Topographic Maps produced by the Department of Survey and Mapping Peninsular Malaysia. The RSO projection (Fig. 2.2.2) was selected because of the shape of the area to be mapped and the scale distortion which can be tolerated.

• WGS (World Geodetic System) 84 coordinates (Jadual 2001 item 1.4 Part I). The world Geodetic System 1984 coordinates are used when Point Positioning is determined using GPS (Global Positioning Satellites). The World Geodetic System (WGS84) the latest revision is WGS84 dating from 1984 (last revised in 2004) will be valid to about 2010. A unified World Geodetic System based on the WGS84 ellipsoid is essential for several reasons:- - International space science and astronautics - Inter-continental geodetic information - Inability of large geodetic systems such as the Rectified Skew Orthomorphic (RSO)

for Peninsular Malaysia which cannot be extended to include islands in the South China Sea nor the East Malaysia State of Sabah and Sarawak; European Datum (ED50) and North American Datum (NAD) to provide worldwide coverage to meet the need for global or regional maps for navigation, aviation and geography

b. Eventually when the GDM2000 coordinates system is fully implemented the requirement for coordinates transformation may be greatly reduced. GDM2000 is described at item 2.2.4.

c. The Consulting Civil and Structural Engineers requirement for engineering survey plans to be in RSO Coordinates against plans in the respective Land Office in Kedah and Pulau Pinang in their respective Cadastral Cassini Solder Coordinates require the mathematical process of transformation of coordinates e.g. WGS84 to RSO or Kedah Cassini to RSO or vice versa

4.14 AIR SURVEY MAPPING TECHNIQUE FOR PRODUCING ENGINEERING SURVEY

PLANS (JADUAL 2001 ITEM 11) The provision here is for the out-put of photo-mosaics and photo-maps over a wide area or long corridor using aerial photographs supplied by the Department of Survey and Mapping. If the Survey Department aerial photographs are out of date “Jadual 2001 item 12” caters for acquisition of new ones by Air Survey methods. The benefits of adopting this approach are:-

a. Access and Coverage - Aerial images can be obtained of areas that are inaccessible or dangerous for ground surveyors due either to unfriendly inhabitant, difficult terrain or a need to maintain confidentiality. An accurate survey can then be compiled in comfortable surroundings. The approximate width of the corridor covered is 1000m (1km) whereas the actual Right of Way (ROW) may be 100m. It provides advance survey information over a wider area which can then be narrowed down to the proposed corridor requiring follow-up of more detailed field survey works.

b. Speed and Cost - Due to the high speed of aerial surveys the cost of works is reduced, and the final product is available earlier. In addition, the expenses of working away from base are reduced, as only the flying crew and some camera operators need travel to the survey area. The photomap together with the photo-mosaic will provide a more focused approach to the planning and scheduling of the actual field survey works.


March 2009 4-7

c. Control - The organization and control of the survey is simplified as the bulk of the surveyors

are working in good stable conditions at base, where they can be easily administered and supervised. Those conditions produce high output and quality of work.

d. Supply - Supply is also simplified as in most cases the aircraft can operate from a commercial airport. Any necessary special equipment such as GPS enabled digital camera can be carried in the survey aircraft to the area of operations.

e. Weather - Although low cloud and extensive cloud-cover will prevent photography, only a short time is needed to obtain suitable images. The weather is therefore seldom a major problem, and once the photographic data has been obtained the survey is unaffected by weather conditions.

4.14.1 Limitation of Air Survey Survey Control - In order to relate an air-survey to the area in which the images were taken, it is necessary to have precise ground coordinates, both plan and height, of points that can be clearly seen on the images and on the ground. Coordinates and a clear description of each point are provided by the ground surveyors as control for the aerial survey. Whilst aerial triangulation using electronic computers provides a means of distributing additional controls on photographs a certain amount of ground control is necessary, and must be provided before the air survey mapping works can be commenced. Invert levels below the water surface cannot be ascertained. Check - A field check of an air survey is necessary to eliminate errors due to misinterpretation of detail. If the survey is at a large scale, completion of hidden detail (under trees, in shadow, etc) may be needed. In all cases, names and description must be obtained from ground survey works. Administrative work include:-

a. Arrangement for tasking of aircraft b. Application for security clearance and the obtaining of the permit to fly aerial photographic

mission c. Mobilization of personnel and equipments

4.15 HYDROGRAPHIC SURVEY FOR TERRITORIAL WATERS AND INLAND WATER

BODIES (JADUAL 2001 ITEM 14 PART V) Hydrographic survey provides information and data to support:-

a. The management of coastal zones b. The hydrographic survey of deltaic regions and river months up to two kilometers upstream

of river mouth c. The development of coastal engineering, property, infrastructure projects and activities d. The management and development of jetties, ports, harbors and associated maritime

facilities e. The management and development along inland waterways and inland water body


4-8 March 2009

4.16 LOCATING OF CROSS-SECTION PROFILES FOR HYDRAULIC ENGINEERING

(JADUAL 2001 ITEM 14.9 PART V) 4.16.1 Mixed Survey Methods Obtaining cross-section profile of stream, adjoining bank and flood plain requires a combination of survey methods. Hydrographic sounding surveys performed in the river must be combined with conventional topographic, and or photogrammetric surveys in the adjacent over banks and flood prone plain. Surveys of the flood plains are usually more efficiently conducted using air survey (Digital photogrammetric) methods to create a Digital Elevation Model (DEM). Recently, airborne LIDAR (Item 4.17) techniques have been developed to provide DEM of the flood plain. Conventional topographic survey methods (leveling and digital/optical total station) will be required to fill in hidden areas under cover of vegetation and to ascertain break lines in the final terrain models. 4.16.2 Guidance to Surveyors on Cross-Section Locations Detailed guidance for determining the location and spacing of stream cross-sections is based on the recommendations in the US Army Corps of Engineers, Engineers Manual “EM1110-2-1002” and EM1110-2-1416”. Surveyors providing input for these studies should be aware of the hydraulic considerations that dictate the intended placement and alignment of stream sections. Thus, knowledge of the engineering rationale for locating cross-sections profiles is required by field surveyors in order to make reasonable adjustments or recommend modification to the project engineer to optimize the obtaining of basic field information on the river profile, the adjoining river banks and the flood plain. 4.16.3 Guidelines on Locating Cross-Sections Generally (not exhaustive) the locations of Cross-sections for hydraulic modeling should be considered are:-

a. Points where roughness changes abruptly to provide channel roughness information b. Closer together in stretches where water surface expands and in bends c. Closer together in stretches where the flow of water changes greatly as a result of changes

in width, depth or roughness d. Closer together at wide bends where the lateral distribution of water flow changes radically

with distance e. Closer together in streams of very low gradient at lowlands which are significantly non

uniform, because the computations are very sensitive to the effects of local disturbances and/or irregularities

f. At tributaries that contribute significantly to the main stem flow. Cross-sections should be located immediately upstream and downstream from the confluence on the main river and immediately upstream on the tributary

g. At regular intervals along waterway of uniform cross-section h. Above, below, and within bridges at bridge sites including the soffit levels i. On large rivers that have average slopes of 0.4 meter to 1.5 meter per kilometer or less,

cross-section within fairly uniform reaches may be taken at intervals of 1.5 km or more j. More closely spaced cross-sections are usually needed to define energy losses in urban

areas, where steep slopes are encountered, and on relatively narrower streams. On small streams with steep slopes it is desirable to take cross-sections at intervals of 500m or less.


March 2009 4-9

k. Recommended maximum reach lengths (distances between cross-sections) are: (1) 800m for

wide flood plains and slope less than 0.4m per km, (2) 550m for slopes less than 0.6m per kilometer, and (3) 365m for slopes greater than 0.6m per kilometer. In addition, no reach between cross-sections should be longer than 75 – 100 times the mean depth for the largest discharge, or about twice the width of the reach. The fall of a reach should be equal to or greater than the largest of 0.15m or the velocity head, unless the bed slope is so flat that the above criterion holds. The reach length should be equal to, or less than, the downstream depth for the smallest discharge divided by the bed slope

Figure 4.2 Typical Cross-Section Configuration

4.16.4 Additional Guidelines on Cross-Section Profiles Field surveyors should also take into consideration the following application when acquiring cross-sectional data.

a. Cross-sections are run perpendicular to the direction of flow at intervals along the river. The “reach length” is the distance between cross-sections. Flow lines are used to determine the cross-section orientation. The hydraulic engineer will provide these orientations to the surveyor.

b. The cross-section should be referenced to the stream thalweg (deepest part of the channel) and by river kilometers measured along the thalweg. From this the reach lengths (distance between cross-sections) is computed. End points on the cross-section should be geographically coordinated using the local State Plane Cassini Soldner Coordinate System.

c. End station elevations. The maximum elevation of each end of a cross-section should be higher than the anticipated maximum water surface level.


4-10 March 2009

d. Local irregularities in bed surface. Local irregularities in the ground surface such as

depressions or rises that are not typical of the reach should not be included in the cross-sectional data.

e. Bent cross-sections. A cross-section should be laid out on a straight line if possible. However, a cross section should be bent if necessary to keep it perpendicular to the expected flow lines.

f. Avoid intersection of cross-sections. Cross-sections must not cross each other. Care must be taken at river bends and tributary junctions to avoid overlap of sections.

g. Inclusion of channel control structures. Channel control structures such as bunds or wing dams should be shown on the cross-section, and allowances in cross-sectional areas and wetted perimeters should be made for these structures.

4.16.5 Cross-Sections Adjacent to Bridges or Culverts (Jadual 2001 Item 3 Part I) Cross-sections need to be denser near bridges and culverts in order to analyze the flow restriction caused by these structures. A guide on the locations of cross-sections is shown below.

Figure 4.3 Cross-Section Locations at a Bridge or Culvert

4.17 LIDAR - LIGHT DETECTION AND RANGING AIRBORNE MAPPING Information on this aspect of surveying, which was described and illustrated in Appendix 3A-2. and in item 3.5 earlier can be found from the web by keying in the following:-

a. LIDAR technologies b. us army corps of engineers hydrographic survey manual (Click item EM1110-2-1003)

LIDAR technology which is similar to radar is an airborne laser mapping technique. A typical airborne LIDAR system is coupled with a Global Positioning System (GPS) to determine aircraft position and an Inertial Navigation System (INS) or Inertial Measuring Unit (IMU) to determine the constantly changing aircraft attitude. Appendix 3A-2 shows the operation of a typical LIDAR using a fixed wing

CROSS SECTION

W

4 X L

CH 001

CH 002

CH 003

CH 004

W L

UP STREAM

DOWN STREAM

BRIDGE/CULVERT

EXPANSION

CONTRACTION

RIVER

L – Length of abutment W- Span of bridge


March 2009 4-11

aircraft. With LIDAR, highly accurate digital elevation (DEM) and digital terrain (DTM) models or elevation contours can be generated immediately well before any aerial photography is processed, ground control is acquired and photogrammetric mapping is performed. LIDAR can capture data with accuracies of 5 to 20 centimeters to meet modeling efforts day and night in a variety of weather conditions. However integrating LIDAR data with photogrammetric data from air survey often yields better end-results since shorelines frequently have heavy ground vegetation cover and mapping goals are frequently 1 to 2 feet (30cm to 60cm) contours. In other words, combined LIDAR – Air Survey/Photogrammetric Mapping provides a more realistic depiction of the terrain and ensures desired map accuracies will be maintained by providing an independent check. Airborne LIDAR system can be broadly classified into 3 main types: wide area mapping systems flown from fixed wing aircraft, Corridor mapping systems from helicopters and Bathymetric mapping systems flown from either one of the platform.


4-12 March 2009

4.18 REFERENCES [1] US Army Corps of Engineers website is accessible by keying in “us army corps of engineers hydrographic survey manual” then click “EM 1110-2-1003 Title: Engineering and Design – Hydrographic Survey” [2] United States Geological Survey website Map Projection Poster egsc.usgs.gov/isb/pubs/MapProjections/projections.html” [3] “The Orthomorphic Projection of the spheroid” Brigadier M. Hotine CBE in the Empire Survey Review vols VIII and IX Nos 62-65, particularly para 19 E.S.R. no. 64 of April 1947


March 2009 4A-1

APPENDIX 4A-1

SCHEDULE ‘C’ – TREASURY APPROVED RATE

(JADUAL FEE UKUR KEJURUTERAAN 2001)


4A-2 March 2009


March 200

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apter 4 REFEREENCES ON GEOOMATICS AND

LAND SURVEY SERVICES

4A--3

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Chaapter 4 REFERE

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March 20009


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March 2009 4A-45

APPENDIX 4A-2

SCHEDULE ‘D’ – AKTA JURUKUR TANAH BERLESEN 1958 P.U. (A) 169.

(Relevant Pages Only)


4A-46 March 2009



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4A-56 March 2009



March 2009 4A-57

APPENDIX 4A-3

MINISTRY OF FINANCE LETTER

ON

MACRES (MALAYSIAN CENTRE FOR REMOTE SENSING)


4A-58 March 2009


March 200

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apter 4 REFERE


4A-5

59


4A-60 March 2009



March 2009 4A-61

APPENDIX 4A-4

BQ EXAMPLE – COST ESTIMATE FOR

SURVEY OF EXISTING ROUTE OF WATERWAYS

CANALS AND DRAINS

(2.4.9)


4A-62 March 2009



March 2009 4A-63

Item Description Estd Actual Unit Rate Agreed Actual Remarks/Qty Qty (RM) Amount Amt Try Item

(a) (b) (c) (d) (e) (f) (g) (h) (I)1 Waterways and hydrographic survey

of Muda River (Tidal River)

1.1 Preparatory work 1 P/day 743.00 743.00 8.1

1.2 Mobilization & demobilization 18 P/day 743.00 13,374.00 8.2(Six Field Parties) respectively

1.3 Planimetric control for as-built 33.8 km 2,500.00 84,500.00 8.3both banks (length 33.8 km)Kedah & Penang

1.4 Height control from existing 13 km 743.00 9,659.00 8.4bench mark (misclosure check)

1.5 Strip survey with details of existing tidal waterway (2 x 250m over banks + 100m waterway )waterway

a) Alignment survey (4.25 x RM243) 13 km 3,157.75 41,050.75 8.11b) Cross-section survey at 100m interval 53.4 km 5,944.00 317,409.60 8.11/3.10.2c) Long-section survey at 13 km 5,944.00 77,272.00 8.11/3.10.2 100m interval

1.6 Establishment of TBMs (Monumentation) 11 No 148.50 1,633.51 8.5

1.7 Site Survey & preparatory works(minimum fee)

a) Site No. 1 min 1 1,486.00 1,486.00 b) Site No. 2 min 1 1,486.00 1,486.00

c) Site No. 3 min 1 1,486.00 1,486.00 7.10 & 8.12d) Site No. 4 min 1 1,486.00 1,486.00 e) Barrage min 1 1,486.00 1,486.00 f) Jambatan Merdeka min 1 1,486.00 1,486.00

1.8 Others

1.9 Re-imbursable cost for purchase 8.8.8.1.10of Revenue Sheet (Std Sheets)CPs, hire of boat and travelling expenses

1.10 Land Acquisition Plans

a) Preparatory work 1 P/day 743.00 743.00 3.11/1.11.1

d) Search at Land Office 16 hour 10.00 160.00 3.11/1.11.3c) Computation Plan 704 lot 20.00 14,080.00 3.11/1.11.4

569,540.86 2 Add 5% Government Service Tax 113,908.17

683,449.03

3 Supply of Land Acquisition PlansPenang 40 sets @ 10 plan/set 40 100 plan 10.00 400.00 1,000.00 8.17/1.14.3

Kedah 40 sets @ 10 plan/set 40 100 plan 10.00 400.00 1,000.00

Estimated Total

Note:a) Alignment survey comprise location of form lines of the waterway

b) Rate of 8 party day per kilometre if the depth of water is more that 1 metre(Specification (viii) Item 8.11 Jadual 2001) for cross-section and Longitudinal section

BQ EXAMPLE - COST ESTIMATE FOR SURVEY OF EXISTINGROUTE OF WATERWAYS CANALS AND DRAINS

APPENDIX 4A-4


4A-64 March 2009



March 2009 4A-65

APPENDIX 4A-5

BQ EXAMPLE – COST ESTIMATE FOR HYDROGRAPHIC SURVEY

OF TERRITORIAL WATERS AND INLAND WATER BODIES

(2.4.16)


4A-66 March 2009



March 2009 4A-67

Item Description Estd Actual Unit Rate Agreed Actual Remarks/Qty Qty (RM) Amount Amt Try Item

(a) (b) (c) (d) (e) (f) (g) (h) (I)

1 Mobilization & demobilization of 3 P/day 743.00 2,229.00 14.1topographic survey equipment

2 Planimetric control and connection in 10 km 2,500.00 25,000.00 14.2built up area

3 Height control and connection 10 km 743.00 7,430.00 14.3

4 Topographic strip survey with details 100 ha 99.00 9,900.00 14.8100m x 10 km coastal strip

5 Bathymetric (Off Shore) Profiling

a) Profiles at 50m 90 km 297.20 26,748.00 14.9.2b) Profiles at more than 100m 30 km 371.50 11,145.00 14.9.2c) Extended hydrographic survey 9 km 222.90 2,006.10 14.9.2 up-stream at 25m intervals

6 Direct Reading of Tide Pole 11 No 148.50 1,633.51 14.10.2

a) Installation of Tide Pole 1 no 900.00 900.00 b) Tidal observation 2 P/day 743.00 1,486.00

88,477.61 7 Add 5% Government Service Tax 17,695.52

106,173.13 8 Boat

a) Mobilization 1 no 600.00 600.00 b) Rental 5 P/day 300.00 1,500.00

Estimated Total 108,273.13

BQ EXAMPLE - COST ESTIMATE FORCOASTAL AND WATERWAYS HYDROGRAPHIC SURVEY

APPENDIX 4A-5


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March 2009 4A-69

APPENDIX 4A-6

TEMPORARY BENCH MARK (TBM)

MARKERS ON NORMAL SURFACE


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March 2009 4A-71

IP. 28 JPS

NAME OF

SURVEYOR

5mm Ø Drilled Center

End Cap Sealed to Pipe

Concrete

50mm Ø G.I Pipe

Figures Engraved on Concrete

300

600

150

300

Proj

ectio

n

TBM MARKER ON NORMAL SURFACE


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March 2009 4A-73

APPENDIX 4A-7

TEMPORARY BENCH MARK (TBM)

MARKERS ON HARD SURFACE


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March 2009 4A-75

Cement/Mortar Mix

10mm Ø Rivet

6mm Thk. Galvanised Steel Plate

30 50 50

50

4 Nos. 150mm Galvanised Steel Nails driven into concrete or hard surface (except pavement)

300

300

30

30

TBM 19

JPSNAME OF SURVEYOR

10mm Ø Rivet Engraved Figures

TBM MARKER ON HARD SURFACE


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CHAPTER 5 GEOGRAPHIC INFORMATION SYSTEM (GIS)

Chapter 5 GEOGRAPHIC INFORMATION SYSTEM (GIS)

March 2009 5-i

Table of Contents Table of Contents .................................................................................................................... 5-i

5.1 INTRODUCTION .......................................................................................................... 5-1

5.2 MORE ON GIS INFORMATION ....................................................................................... 5-1

5.3 REFERENCES ............................................................................................................... 5-2


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March 2009 5-1

5 GEOGRAPHIC INFORMATION SYSTEM (GIS)

5.1 INTRODUCTION There is an old common saying that “a picture is worth a thousand words”. Today with GIS we can choose to have not only the picture which is very often the digital topographic map but also the thousand words which is interlinked with the geographically referenced features depicted on a map through feature codes. A GIS is a computerized system capable of capturing, storing, analyzing and displaying geographically referenced information; that is data of map features identified according to location. Traditionally such a graphic picture is depicted on cartographically enhanced topographic maps (USGS website on geographic information system http//:egsc.usgs.gov/isb/pubs/gis_poster/). GIS tools and methods can be used for environmental studies, water resource management for agriculture, flood mitigation development planning or scientific investigation. A GIS may allow flood emergency planners to easily calculate flood emergency response times during a flood season. Together with cartography a component of topographic mapping, remote sensing, global positioning systems, photogrammetry, and geography; GIS has evolved into a discipline with its own research base known as Gographic Information science An example on the usefulness of GIS technology development is the possibility of combining agricultural or land records, hydrography; which include rainfall data, to determine which river will carry certain levels of soil erosion sediment runoff. Having gone through the above it is hoped the user of this manual can now make use of the link provided by the Malaysian Centre for Geospatial Data Infrastructure [2] (MaCGDI) Ministry of Natural Resources and Environment (NRE) website http://www.mygeoportal.gov.my to contact various other departments to share experience and ideas on creating geospatial information. 5.2 MORE ON GIS INFORMATION More information which is listed below can be obtained from the USGS website mentioned in item 5.3 References.

• How does a GIS work? • Data Capture • Data integration • Map projection and registration • Data structures • Data modeling • What’s special about a GIS? • Framework for cooperation etc.


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5.3 REFERENCES [1] Malaysian Centre for Geospatial Data Infrastructure (MaGDI) website http://www.mygeoportal.gov.my

CHAPTER 6 CHECKLIST FOR TERRAIN FEATURES


March 2009 6-i

Table of Contents


6.1 SURVEY SERVICES ....................................................................................................... 6-1

6.2 LAND ACQUISITION BASE PLAN. .................................................................................. 6-1

6.3 GROUND MARKERS ...................................................................................................... 6-1

6.4 INDUSTRY .................................................................................................................. 6-1

6.5 ROAD FURNITURE, SERVICES AND UTILITIES ............................................................... 6-2

6.6 BOUNDARY FEATURES ................................................................................................. 6-2

6.7 BRIDGE SITE ............................................................................................................... 6-2

6.8 RAILWAYS .................................................................................................................. 6-2

6.9 SURVEY CONTROL ....................................................................................................... 6-3

6.10 PLANTATIONS, TREES AND RECREATIONAL AREAS ........................................................ 6-3

6.11 SLOPES AND EARTHWORKS ......................................................................................... 6-3

6.12 WATER AND DRAINAGE ............................................................................................... 6-3

6.13 REFERENCES ............................................................................................................... 6-4


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(This page was intentionally left blank.)


March 2009 6-1

6 CHECKLIST FOR TERRAIN FEATURES

6.1 SURVEY SERVICES The survey services to be provided by the surveyor are as listed below and as detailed: -

a. Discussion with relevant authorities such as JKR, JPS, Survey Department, Local Authority and Land Office before the physical commencement of work on site

b. Consultation with the Superintending Officer (SO) or SO’s Representative and obtain Instructions

c. Study all relevant information, maps and plans provided and obtaining all necessary additional topographic maps, certified plans, revenue sheets, data and other information for the proper execution of the works

d. Preparation of topographic survey plans e. Field survey to pick up details according to format required f. Compiling, processing and preparing data and CAD plot of survey plan in accordance to

format required g. In carrying the work, the surveyor shall attempt to obtain permission prior to entry into

private land, cemeteries and property of other relevant authorities 6.2 LAND ACQUISITION BASE PLAN. The drawing shall show the following:

a. Name of districts and mukims b. Lot boundaries and lot numbers c. Existing total lot areas computed based on coordinates d. Land use indicating type of cultivation etc. e. Type of building indicating permanent or semi permanent and usage f. The existence of burial ground if any within the survey corridor g. All other relevant details as instructed by client or as desired by the government h. Land lots that are partially within the mapping area shall, where possible, be presented

showing the whole area of the lot 6.3 GROUND MARKERS The surveyor shall supply two copies of the following results to the client on completion of field work and adjustment:

a. Schedule of all Permanent Ground Markers (TBM’s and RM’s) giving the reference numbers, coordinates and heights

b. Descriptions of Permanent Ground Markers giving the types of marker constructed and location

c. Diagrams of the horizontal control net showing the connection between Permanent Ground Markers

d. Diagrams of the leveling (height control) net indicating the connection between Permanent Bench Marks

6.4 INDUSTRY

a. Tanks b. Valve chambers c. Transformers (boundary fences and building lines) d. Electricity sub-station, boxes and switch boxes (boundary fences and building lines) e. Pylon lines (indicate levels at lowest point at sag and at pylon towers) f. Pylon bases g. Pylon reference numbers and


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h. Telegraph lines

6.5 ROAD FURNITURE, SERVICES AND UTILITIES a. Km post (value to be noted) b. Guardrails c. Bus stops d. Lamp posts e. Telecom poles f. Electricity poles g. Road signs h. Large road signs (with minimum 2 posts only) i. Hoardings j. Large notice boards and display boards k. Traffic signals and control boxes l. Vehicle detector pads m. Road drains or gullies n. Fire hydrants o. Stop valve and stand pipes p. Top of manholes (circular and square) q. Weigh bridge; and r. Services above ground (such as some water pipelines) 6.6 BOUNDARY FEATURES a. Fences b. Gates c. Hedges d. Walls e. Burial grounds (indicate whether Muslim,. Chinese, Christian etc.) and f. Historical areas 6.7 BRIDGE SITE a. Width of bridges b. Soffit levels of edge beam c. Carriage way d. Existing reserve e. Size, type and location of utility services adjacent and along the span of the bridge f. Spans and location of columns/piers g. Level of water and date taken 6.8 RAILWAYS a. Railway running rails b. Points c. Bridges (over roads, river, etc.) d. Signal boxes e. Telephone points f. Telegraph poles, and g. Km posts (value to be noted)


March 2009 6-3

6.9 SURVEY CONTROL a. Survey Department GPS and Boundary Marks for horizontal control b. Ground control points c. Permanent ground control markers d. Survey Department Bench Marks (BM) vertical control; and e. Temporary Bench Mark (TBM) established 6.10 PLANTATIONS, TREES AND RECREATIONAL AREAS a. Playing fields b. Parks and open spaces c. Laid out pitches d. Prominent trees; and e. Land-use and vegetation etc. 6.11 SLOPES AND EARTHWORKS a. Cutting and embankments b. Terraced slope c. Ornamental slopes d. Mounds e. Industrial waste; and f. Refuse tips 6.12 WATER AND DRAINAGE a. Rivers (name to be indicated) b. Streams c. Water courses d. Ditches (width and depth to be indicated) e. Swamps f. Lined drains (type, size, depth to be indicated) g. Culverts with sizes and invert levels, including sketch of inlet and outlet structures such as wing

wall h. Irrigation structures such as Weirs, bunds, spillways, barrage, floodgates, dams and floodwalls i. Pump station sites j. Tanks k. Sewer outfalls and top of manhole covers l. The top of all water features over 1.0 meter wide are to be detailed and the bottom of banks as

indicated by the water level at the time of the survey. The direction of flow of all rivers, streams and watercourses is to be indicated

m. Slopes with a height greater than 1.0 meter or too sharp a gradient to be shown by contours, including river banks, are to be shown by conventional markings and the top and bottom of slopes are to be shown as dotted lines; and

n. Slope conventions are to be drawn as near as possible to indicate the actual shape of the slope face, i.e. all berms and terraces are to be detailed

o. Flood spillways and closure bunds p. Tidal variation sites for tidal gate structures or bunds q. Highest known flood level Any other visible features not listed likely to affect design and later construction works are also to be shown.


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REFERENCES [1] Department of Survey and Mapping website http://www.jupem.gov.my [2] Malaysian Centre for Geospatial Data Infrastructure (MaGDI) website http://www.mygeoportal.gov.my [3] Digital Globe for Satellite Imagery at website http://www.digitalglobe.com [4] US Army Corps of Engineers website is accessible by keying in “us army corps of engineers hydrographic survey manual” then click “EM 1110-2-1003 Title: Engineering and Design – Hydrographic Survey” [5] United States Geological Survey website Map Projection Poster egsc.usgs.gov/isb/pubs/MapProjections/projections.html” [6] “The Orthomorphic Projection of the spheroid” Brigadier M. Hotine CBE in the Empire Survey Review vols VIII and IX Nos 62-65, particularly para 19 E.S.R. no. 64 of April 1947 [7] GDM2000 Geodesy Section, Department of Survey and Mapping website http://geodesi.jupem.gov.my

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