Chapter 2

Review of literature

2.1 INTRODUCTION

Expansive soils are highly problematic as they have a tendency to increase in

volume on absorption of water and to shrink on evaporation of water. On absorption of

water the density of these soils decreases and they become slushy; they become hard on

evaporation of water due to increase in density. The volumetric change in these soils is

attributed to seasonal variations in the ground water profile resulting in changes in the

moisture content (Rees and Thomas, 1993).

The entire stratum of expansive soil in the field may not be active, as most soils

do not respond quickly to climatic changes. During summer, polygonal shrinkage cracks

appear on the surface, extending to a depth of about 2m, indicating a high potential for

expansion and contraction (Mohan, 1977). The depth of cracking indicates the depth of

active zone in which significant volume changes occur because of high moisture

deficiency (Snethen, 1980).

Katti (1978) observed that, in Indian expansive soils, the depth of active zone is

confined only to the top 1.0 to 1.2m. According to Mohan (1977), however, significant

ground movements occur even up to a depth of 3.5m below the ground level.

Ramaswamy (1990) observed that, in certain areas, the depth of active zone might extend

beyond 3.5m also.

2.2 IDENTIFICATION OF EXPANSIVE SOILS Expansive soils can be identified by their mineralogical composition, index

properties, suction movements or direct measurement of swelling characteristics.

2.2.1 MINERALOGICAL COMPOSITION

The amount of volume change by an expansive soil depends greatly on its

mineralogical composition. Expansive soils are usually composed of montmorillonite

clay minerals (Omari and Hamodi, 1991). The presence of montmorillonite, which has

4

an expanding lattice, contributes most to swelling, clearly indicating the expansiveness

of a clay soil (Shreiner, 1987).

The various techniques employed to determine the mineralogical composition are

X-ray diffraction, differential thermal analysis (DTA) and electron microscope resolution

(EMR). The X-ray diffraction technique gives the proportions of various minerals

present in a colloidal clay fraction. The DTA method is based on the fact that certain

characteristic reactions take place at specific temperatures for different minerals when

these minerals are heated to high temperatures, resulting in a loss or gain in heat. EMR

helps to determine the texture and internal structure of the mineral (Chen, 1988).

2.2.2 IDENTIFICATION FROM INDEX PROPERTIES

The degree of swelling of an expansive soil can be estimated through index

properties also. The Atterberg limits, clay content, activity etc. are used as parameters to

categorize the degree of swelling (Chen, 1988).

Skempton (1953) observed that volume change characteristics of a soil depends

on “activity”, which is defined as the ratio of the plasticity index of the soil to the

percentage clay particles i.e., fraction of soil finer than 2 micron (0.002mm).

Subsequently, this definition was modified for natural soils (Seed et al. 1962) as

5% - m2 finer thanclay %

Index µ

PlasticityActivity= (2.1)

A soil with an activity of less than 0.75 is termed as inactive i.e., it indicates a

low potential for volume change and a soil with an activity of more than 1.25 is termed

as active, indicating a high potential for volume change. The values lying between 0.75

and 1.25 indicate medium potential and soils with these values are classified as normal.

Altmeyer (1955) characterized the degree of expansion of a soil based on its

shrinkage limit and linear shrinkage as given in Table 2.1. Linear shrinkage is defined as

the ratio of decrease in linear dimension of a specimen to its initial dimension, expressed

as a percentage and given by the nearest whole number.

5

Table 2.1 Degree of expansion based on linear shrinkage

Shrinkage limit

(%)

Linear shrinkage

(%)

Probable swell

(%)

Degree of expansion

< 10 >8 >1.5 Critical

10 – 12 5 – 8 0.5 – 1.5 Marginal

> 12 < 5 < 0.5 Non – critical

Holtz and Gibbs (1956) developed a simple test called the free swell test for the

determination of swell potential of a soil. The test is performed by pouring 10cc of dry

soil passing 425micron sieve into a graduated cylindrical glass jar of 100ml capacity

filled with water and observing the swollen volume. Free swell index (FSI) is expressed

as a percentage increase in the volume to the original volume of the soil. Soils having a

free swell of more than 100% pose more damage to lightly loaded structures, and those

having an FSI of less than 50% do not pose serious problems to structures.

Mohan and Goel (1959) suggested a more convenient method, which was later

adopted by Indian Standards Institution. In this method, 10g of oven dried soil passing

425 micron sieve is poured separately into two graduated cylindrical glass jars of 100ml

capacity, one containing distilled water and the other kerosene. Allowing both the jars to

stand for 24 hours, the final volumes of the soil in the two cylinders are noted. The

differential free swell, which was later redesignated as free swell index (FSI), expressed

as a percentage, is given as,

(2.2) 100×

−=

kVks VVFSI

where, Vk = final volume of the soil in kerosene, which is a non-polar liquid, and

Vs = final volume of the soil in water which causes swelling of the soil.

According to Mohan and Goel (1959), a soil with an FSI of 50% or more has a

very high swell potential, and that with an FSI of 20% or less has a low swell potential.

Holtz (1959) proposed some identification criteria based on the probable volume change

in expansive soils, which is the percentage change in the thickness of the sample from

6

oedometer swell test under a surcharge of 6.9kPa or 1psi from the air-dry condition to

saturation. To calculate the probable expansion, the three properties of soil given in

Table 2.2, in which the classification is also given, should be considered simultaneously.

Table 2.2 Degree of expansion based on PI and SL

Colloidal

content %

Plasticity index

(%)

Shrinkage limit

(%)

Probable volume

change (%)

Degree of

expansion

>28 >35 <11 >30 Very high

20 – 31 25 – 41 7 – 12 20 – 30 High

13 – 23 15 – 28 10 – 16 10 – 20 Medium

5 < 18 > 15 < 10 Low

Seed et al. (1962) proposed a relationship between swell potential (S.P) and

plasticity index (PI) as,

S.P = 2.16 X10 -3 (PI) 2.44 (2.3)

where swell potential is the ratio of the increase in thickness to the original thickness of

the soil sample in a consolidation ring, compacted at optimum moisture content, soaked

in water under a surcharge of 6.9 kPa (1Psi). This is expressed as a percentage.

Seed et al (1962) gave an expression for swell potential (SP) in terms of activity (A) and

clay content (C) as

S.P = 3.6 X10-5 (A 2.44) (C 3.44) (2.4)

Seed et al (1962) gave another expression also for swell potential (S.P) in terms

of percent clay (C) i.e., soil fraction with size less than 2 micron as

S.P = k C x (2.5)

where ‘x’ is an exponent and ‘k’ is a coefficient, both depending upon the type of clay.

Ranganatham and Satyanarayana (1965) proposed a relationship between swell

potential (S.P) and shrinkage index (S.I) as

S.P= β (S.I)z (2.6)

β = 1/6.3 and z = 1.17 for natural soils and

β = 1/256 and z = 2.37 for artificially prepared sand – clay mixtures.

Based on this relationship, the degree of expansion is classified as shown in Table 2.3.

7

Table 2.3 Degree of expansion based on shrinkage index (Ranganatham and Satyanarayana, 1965) Shrinkage index (%) Swell potential (%) under a

surcharge of 6.9kPa

Degree of expansion

< 20 <10 Low

20 -30 10 – 20 Medium

30 – 60 20 – 30 High

> 60 > 30 Very high

Peck et al. (1974) have related plasticity index to the swelling potential, as shown

in Table 2.4.

Table 2.4 Characterisation of swell potential based on PI

Swelling potential Plasticity index (%)

Low 0 – 15

Medium 10 – 35

High 20 – 35

Very high 35 and above

Chen (1975) gave a relationship for the percentage swell (S) of undisturbed soils

in terms of plasticity index (P.I) in the form

S = B. e A (P.I) (2.7)

where A and B are constants equal to 0.0838 and 0.2558 respectively.

The surcharge pressure used was 6.9kPa. The water content varied between 15% and

20 % and the dry density between 1.6 Mg/cu.m and 1.76 Mg/cu.m.

Mohan (1977) categorized degree of expansion based on liquid limit, plasticity

index, shrinkage index and colloid content as shown in Table 2.5.

8

Table 2.5 Degree of expansion based on liquid limit, plasticity index, shrinkage index and colloid content (Mohan, 1977). Liquid limit

PI Shrinkage index

Colloid content

Degree of expansion

Degree of severity

70 -90 >32 >60 >28 Very high Very critical

50 – 70 23 – 32 30 – 60 20 - 32 High Critical

35 – 50 12 – 23 15 – 30 13 - 23 Medium Marginal

20 – 35 12 <15 <15 Low Non- critical

Shrinkage index in the above table is defined as the difference between liquid

limit and shrinkage limit, while colloid content represents the soil fraction finer than

0.001mm.

Murthy and Raman (1977) defined a term “swelling index” (Isw) expressed as a

fraction and given in the form

Isw = SI/(100+ ws) (2.8)

where ‘SI’ is the shrinkage index (%) and ws is the shrinkage limit (%). Based on

shrinkage index, the degree of expansiveness is classified as shown in Table 2.6.

Table 2.6 Characterization of degree of expansiveness based on swelling index

Swelling index (Isw) Degree of expansiveness

0 -0.1 Low

0.1 – 0.2 Medium

0.2 – 0.5 High

0.5 – 0.75 Very high

0.75 – 1.0 Extra high

Williams and Donaldson (1980) classified the degree of expansiveness

considering the plasticity index of the whole sample, PIws, which is given as

sample whole theof mass

sieve same thefiner than soil of mass 425 through passingfraction soil of P.I. P.Iws ×= µ (2.9)

9

The classification is shown in Table 2.7.

Table 2.7 Degree of expansiveness based on plasticity index of whole sample

(Williams and Donaldson, 1980)

P.I of whole sample Degree of expansiveness

>32 Very high

24 – 32 High

12 – 24 Medium

<12 Low

Bandyopadhyay (1981) proposed an equation for swell potential (S) in terms of

colloidal content (C) as,

S =k’ C x (2.10)

where k’ = a.Ab

in which ‘A’ is the activity of the soil and ‘x’ is an exponent equal to 3.44. The values of

‘a’ and ‘b’ depend upon the plasticity index of the soil. The average values of ‘a’ and ‘b’

applicable to all soils are, a = 2.2x10-5 and b = 2.14

The Indian Standard Institute, now called the Bureau of Indian Standards, has

adopted the classification criteria suggested by Mohan (1977) with a modification

according to which the colloidal content is replaced with the free swell index, as shown

in Table 2.8.

Table 2.8 Characterisation of degree of severity

Liquid limit (%)

Plasticity index (%)

Shrinkage index (%)

Free swell index (%)

Degree of expansion

Degree of severity

70 – 90 >32 >60 >200 Very high Severe

50 – 70 23 – 32 30 – 60 100 – 200 High Critical

35 – 50 12 – 23 15 – 30 50 – 100 Medium Marginal

20 – 35 <12 <15 <50 Low Non-critical

10

2.2.3 SOIL SUCTION

Soil suction which indicates the affinity of soil for water is one of the intrinsic

properties that characterizes an expansive soil and indicates the affinity of soil with water

(Snethen, 1979). It is a measure of the deficiency of moisture in a soil and its capacity to

attract water into its pores.

The total soil suction is a sum of the matrix suction or capillary potential and the

osmotic suction. The capillary potential is equivalent to the negative pore pressure and is

often identified in terms of suction in an unsaturated soil. Volumetric changes respond to

changes in capillary potential (Reese and Thomas, 1993).

Because of this negative pore pressure, expansive soils have remarkable affinity

for water. Hence, soil suction is a fundamental parameter to evaluate wetting (Kenneth et

al. 1993). However, in soils compacted at a water content below plastic limit, suction has

no significant effect on the rate of advance of wetting front. At a water content higher

than plastic limit, suction has a control over the rate of wetting (Weisberg et al. 1990).

Several researchers (Richards, 1967; Aitchison, 1973; Brackley, 1980; Johnson,

1981; Mc Keen, 1981; Snethen, 1980; Richards et al. 1983; Yoshida et al. 1983)

evaluated swell based on the relationships between suction and water content. Swelling

is linearly proportional to the gravimetric water content (Dhowian, 1990). Dhowian

(1992) developed suction potential model to estimate suction as a function of time,

swelling and depth.

Suction can be measured by the filter-paper method (Mc Keen, 1988), in which

the soil sample placed in the presence of a filter paper is allowed to equilibrate and the

water content of the filter paper measured. The suction of the soil sample can be obtained

from a calibration curve for the filter paper. Suction depends on the voids ratio in the

range of low values of suction (Keissar et al. 1990).

2.2.4 PREDICTION OF SWELLING CHARACTERISTICS BASED ON

PLACEMENT CONDITIONS

Placement conditions, which include surcharge, water content and dry density of

the soil, have bearing on the percentage swell and the swelling pressure. The effect of

different placement conditions is explained below:

11

2.2.4.1 Surcharge load and stress history: The increase in surcharge load on an

expansive soil reduces swell. For in situ conditions, any form of overburden on the soil

will reduce swell. Swelling pressure measured depends on the initial surcharge applied

on the specimen (Satyanarayana, 1966). But, according to Chen (1988), it is independent

of the surcharge.

2.2.4.2 Initial dry density and water content: Both swell potential and swelling

pressure increase with increase in dry density (Ranganatham and Satyanarayana, 1965).

According to Satyanarayana, (1966) and Vijayvergiya and Ghazzaly (1973), the initial

moisture content affects swelling pressure whereas, according to Chen (1988) and

Brackley (1973), it does not. The lower the initial water content, the higher will be the

suction and the consequent volume change.

2.2.4.3 Method of compaction: Statically compacted specimens swell more than those

compacted by kneading (Parcher and Liu, 1965). Some more factors that influence the

swelling characteristics are side friction (EI Sayed and Rabbaa, 1986), size of the

specimen (Uppal and Palit, 1969; EI Sayed and Rabbaa, 1986) and deformability of

oedometer (Fredlund, 1982; EI Sohby et al. 1989).

2.2.5 DIRECT MEASUREMENT OF PERCENT SWELL

The most convenient method of determining the swelling characteristics of

expansive clays is by direct measurement in the laboratory.

For the determination of percent swell, the soil specimen is sandwiched between two

porous stones and confined in an oedometer ring. The sample is inundated with water.

Water enters both from the top and the bottom. The final increase in the thickness of the

sample following absorption of water by the soil specimen is recorded and reported as a

percent of the original thickness.

2.2.6 DIRECT MEASUREMENT OF SWELLING PRESSURE

Swelling pressure of an expansive soil is determined by the following three methods in

the laboratory:

i) the method of different surcharges,

ii) the constant volume method and

12

iii) the free swell method ( or, the swell – consolidation method).

Based on the method of determination, swelling pressure is defined (Jennings, 1963) as,

i) the pressure corresponding to zero volume change obtained by interpolation

from the equilibrium swell or compression values of identical specimens

inundated under different surcharges (known as the method of different

surcharges). The swelling pressure thus determined is designated by the

symbol, psds;

ii) the pressure that is necessary to hold the soil sample at its original volume

when it is inundated under no – load condition, or under a surcharge equal to

the field effective overburden pressure (known as the constant volume

method). The swelling pressure thus determined is designated by the symbol,

pscv and

iii) the pressure that is required to recompress a fully swollen soil sample (on

inundation) to its original volume under a nominal surcharge of about 10kPa

(known as the free swell method). The swelling pressure thus determined is

designated by the symbol psfs.

Of these, the free swell method has been found to give the largest value of swell

(Shanker et al. 1982; Sridharan et al. 1986). However, while the former investigators

found the constant volume method to yield the least value, the latter found the method of

different surcharges to yield the least value.

Sullivan and Mc Clelland (1969) and Uppal and Palit (1969) recommend the use of

the constant volume method as it simulates the stress path followed in the field.

However, Holtz and Gibbs (1956) maintain that the entry of water into the sample under

load is difficult as the fissures would be closed. This affects the determined value of the

swelling pressure. Zeitlen (1961) and Justo and Saetersdal (1979) recommend the use of

different surcharges method as it gives the swell expected under any initial surcharge.

Jennings (1965), however, observed that the stress path is not an important factor

in the determination of the percent swell, even though Brackley (1975) and Sullivan and

Mc Clelland (1969) emphasize its importance.

13

2.3. EXISTING FOUNDATION PRACTICES IN EXPANSIVE SOILS The various options in foundation practices adopted to minimize heave in

expansive soils are,

a) avoiding expansive material,

b) mechanical, physical or chemical alteration and

c) adopting special foundation techniques.

2.3.1 AVOIDING EXPANSIVE MATERIAL

Avoiding the expansive soil in favour of a safer foundation soil is not an

economically viable proposition in most of the situations. With the development of

several modern techniques for effectively combating problems posed by expansive soils,

it is seldom adopted these days.

2.3.2. ALTERATIONS FOR MINIMIZING HEAVE

2.3.2.1. MECHANICAL ALTERATIONS

These include excavation of expansive soil and replacement with non-expansive

material, where the depth of active zone is small and where a suitable replacement

material is available. Sand cushion method (Satyanarayana, 1966) and cohesive non-

swelling (CNS) layer method (Katti, 1978) are very popular.

2.3.2.1.1. Sand cushion method: In this method, the entire depth of the expansive clay

stratum if it is thin, or a part thereof, if it is deep enough, is removed and replaced by a

sand cushion compacted to the desired density and thickness. Swelling pressure varies

inversely as the thickness of the sand layer and directly as its density. Hence, sand

cushions are formed in their loosest state without violating the criterion of bearing

capacity. Fig 2.1 shows the concept of sand cushion technique.

Expansive clay

Expansive clay

Sand cushion

Fig 2.1 Sand cushion method

14

The basic philosophy of this method is that, in monsoon, the saturated sand occupies less

volume, accommodating some of the heave of underlying soil, and in summer, partially

saturated sand bulks and occupies the extra space left by the shrinkage of the soil.

This method, however, has some limitations. Firstly, the high permeability of sand

creates accumulation of water, and secondly, the thickness of the sand cushion depends

on the depth of active zone termine. Besides, effects of

tage with this system is, it is not easy to get a material which conforms to the

specifica

pansive soil were arbitrarily

and 150-75 µm,

sed on the grain size. Swell potential and

swelling p

increasing fines content. Coefficient of

which itself is difficult to de

fatigue in swelling have to be considered in designing the thickness of the sand cushion.

2.3.2.1.2. Cohesive non- swelling (CNS) layer method: In this method, about top 1m

to 1.2m of the expansive soil is removed and replaced by a cohesive non-swelling soil

layer.

According to Katti, with saturation of expansive soil, cohesive forces are

developed up to a depth of about 1.0m – 1.2m and counteract heave. The electric charge

of clay particles produces absorbed water bonds and develops this cohesion. A CNS

layer creates an environment similar to that around 1m depth in an expansive soil with

equivalent cohesion to counteract heave. Moorum is an example of CNS material. The

disadvan

tions laid down by Katti for an ideal CNS material.

2.3.2.2. PHYSICAL ALTERATION

In this, granular material is mixed with expansive clay to minimize swelling

properties (Satyanarayana, 1966). Phanikumar et al. (2012) studied swell-consolidation

characteristics of these artificially prepared sand-clay mixes in one dimensional

consolidometer. Fine sand content and fines content in the ex

varied in the investigation. The fines content was varied as 425-300 µm

separated from the same expansive soil ba

ressure decreased with increasing fine sand content but increased with

compressibility, coefficient of volume

compressibility and compression index of the samples decreased initially up to a sand

content of 15% and thereafter increased at higher sand contents. One of the

disadvantages of clay-sand and clay-gravel mixes is the faster ingress of water due to

increased permeability.

15

2.3.2.3. CHEMICAL ALTERATION

This involves addition of chemicals to expansive clay to reduce heave by altering

the nature of clay minerals. Stabilisation of expansive soil with various additives

including lime, cement, calcium chloride and fly ash has shown a promising reduction in

heave and improved strength characteristics. (Shanker and Maruthi, 1989; Desai and

Oza, 1997; Phanikumar et al. 2001; Cokca, 2001).

cur when fly ash is blended with expansive soils

(Cokca

would decrease causing reduction in

free sw

and an increase in maximum dry density

and she

Two mechanisms are likely to oc

, 2001; Phanikumar and Sharma, 2007). They are (i) physico-chemical

interactions and (ii) mechanical changes. Expansive clay particles are replaced by the

non plastic fines of fly ash. Fly ash, which is composed of silica, alumina and iron

oxides, promotes flocculation of clay particles by cation exchange. The surface area of

the flocculated particles and their affinity for water

ell index, swell potential and swelling pressure. The reduction in swelling

pressure on addition of fly ash is due to the reduction in the amount of suction (Sharma,

1998). Swell potential and swelling pressure decreased by nearly 50% at 20% fly ash

content (Phanikumar and Sharma, 2004). It was also observed that swell pressure

reduces with increase in the curing time (Cokca, 1999). Further, Phanikumar et al. (2009)

developed the technique of compacted fly ash columns (FAC) as an innovative

foundation technique for expansive clay beds.

Phanikumar (2009) compared the effect of lime and fly ash on free swell index,

swell potential, swelling pressure, co-efficient of consolidation, compression index,

secondary consolidation characteristics and shear strength. Lime content was varied as

2%, 4% and 6% whereas fly ash content was varied as 0%, 10% and 20%. It was noted

that at 20% fly ash there was a significant reduction in swell potential, swelling pressure,

compression index and secondary consolidation

ar strength. A similar pattern was observed when the soil was stabilized with 4%

lime.

A technique called lime-slurry pressure injection (LSPI) is used wherein lime

slurry is injected into drill- holes under a pressure of 15kg/cm2. Lime columns or lime-

soil columns were also tried to stabilize expansive clays in-situ (Rao, 1984). It has been

16

reported (Venkataratnam et al. 1985; Shanker and Maruthi, 1989) that diffusion of lime

is effective up to a radial distance of about 3 times the diameter of the lime soil column.

Table 2

ansive soil (Rao et al. 2008). In this technique fly ash cushion was stabilized

with 1

% of lime.

ent of 8%.

e in the thickness of the

cushion

.9 shows the summary of the data on stabilizing expansive soils with different

chemicals.

Lime-stabilized fly ash cushion ( Rao et al. 2008), stabilized with 10% lime and

with thickness equal to half that of the active zone, found that heave reduced by about

60% and further reduced to 99.2% in the subsequent swelling and shrinkage cycles.

Cement stabilized fly ash cushion technique was also found to be effective in controlling

heave of exp

0% cement with the thickness equal to that of the expansive clay bed. 75% of

heave was reduced when compared with CNS cushion. Heave reduced to 99.1% at the

end of the fourth swell-shrink cycle.

Effect of lime, cement and artificial pozzolan and the combination of these three

stabilizers at different proportions were studied (Al Rawas et al. 2005). Lime, cement

and sarooj (artificial pozzolan produced by burning calcining clay) were mixed at

dosages of 3%, 6% and 9% by dry weight of soil. It was observed that swell percentage

and swell pressure reduced to zero at 6

Sahoo and Pradhan (2010) studied the effect of lime stabilized soil cushion on

strength behaviour of expansive soil. In this study, lime contents of 2%, 4%, 6%, 8% and

10% by dry weight of cohesive non swelling soil were used. Tests were conducted at

different soaking periods of 7, 14, 28 and 56 days. Maximum increase in strength was

observed at 14 days of curing period at a lime cont

Rice husk ash stabilized with lime or cement was used as cushion between the

expansive soil and the foundation to counteract the effect of heaving (Sivapullaiah et al.

2004; Sharma et al. 2008). Rice husk ash was stabilized with a lime content of 3% to

9% or 10% of cement and cured for a period of 7 days to effectively function as a

cushion. It was observed that heave decreased with increas

. Lime stabilized cushion was found to be more effective when compared to rice

husk stabilized with cement.

17

Table 2.9 Summary of data on stabilizing expansive soils with different chemicals

Type of stabilization

Properties of stabilizers

% of stabilizer

Tests conducted Conclusions References

Lime

Hydrated

lime

Lime-slurry

2% -8%

2500 lbs of

lime:900

gallons of

water

Consistency limits, Proctor

compaction, free swell and

shear strength tests

Lime-slurry pressure

injection (LSPI)

Liquid limit, plasticity index and swell

potential of expansive clays decrease

with increase in lime content up to 4%.

Plastic limit and shear strength increase.

MDD increases and OMC decreases.

Diffusion of lime is effective up to a

radial distance of 3 times the diameter

of the lime soil column

Abduljawad

(1995)

Basma et al.

(1998)

Rao (1984)

18

Fly ash

Class C fly

ash

Class C fly

ash

Class C fly

ash

Class C fly

ash

0%, 3%,

5%, 8%,

10%,15%,

20% and

25%

15% and

25%

0%, 5%,

10%, 15%

and 20%

0%, 5%,

10%, 15%

and 20%

Liquid limit, plastic limit

and swell pressure

Plasticity index, swell

potential and cation

exchange capacity(CEC)

Free swell index (FSI),

swell potential, swelling

pressure, compression

index (Cc) and secondary

consolidation

Plasticity index, hydraulic

conductivity, compaction

and undrained cohesion

Grain size distribution of the soils is

altered by the addition of fly ash.

Liquid limit and plasticity index

decreased with increasing fly ash

content.

CEC reduces due to the formation of

new pazzolanic reactions causes the soil

to become more granular and results in

less water absorption potential.

20% of fly ash content reduced FSI,

swelling pressure and swell potential by

about 50%. Cc and secondary

consolidation decreased by 40% at 20%

fly ash.

Plasticity index reduced by 50% at 20%

fly ash.Hydraulic conductivity and

OMC decreased whereas MDD and

undrained cohesion increased.

Cokca (1999)

Cokca (2001)

Phanikumar and

Sharma (2007)

Phanikumar and

Sharma (2004)

19

Class F fly

ash

0%, 3%,

6%, 9%,

12% and

15%

Swell potential, swelling

pressure and unconfined

compressive strength

(UCS)

Swell potential and swelling pressure

decreased with increase in curing time.

UCS increased with increase in curing

time.

Optimum content of fly ash was found

to be between 9% and 12% at a curing

time of 7 days.

Fesheng Zha et

al. (2008)

Cement Portland

cement

Cement kiln

dust

3% -9%

0%-30%

Consistency limits, swell

potential and shear strength

Plasticity index, free swell

and unconfined

compressive strength

Reduces liquid limit, plasticity index

and swelling potential, and increases the

shrinkage limit and shear strength

Reduces plasticity index and swell %

and increases compressive strength

Nelson and

Miller (1992)

Basma et al.

(1998)

Zaman et al.

(1992)

Calcium

chloride

Powder form 1% Plasticity index and swell

pressure

Plasticity index and swelling pressure

decreased with increasing CaCl2

content.

Ramanamurty

and Harikrishna

(2006)

20

Expansive soil can also be stabilized with calcium chloride also (Ramanamurthy

and Harikrishna, 2006). One percent of calcium chloride was applied to the expansive

clay bed by ponding and also through boreholes, and plasticity of the clay reduced to

nearly 40% - 60% whereas the plasticity reduced to only 7% -15% when lime slurry was

used. Similarly the swell pressure of the clay decreased by 50% – 60% and 20% - 25%

respectively in cases of CaCl2 application and lime application. It was observed that

calcium chloride was several times better than conventional lime.

2.3.2.4. FIBER REINFORCEMENT OF EXPANSIVE SOILS

Fiber reinforcement of expansive soils also proved to be quite successful in

reducing volume changes and increasing shear strength of expansive soils. Puppala and

Musenda (2000) investigated the effect of discrete and randomly oriented polypropylene

fibre reinforcement on strength and volume change behaviour of expansive soils. Fiber

reinforced clayey samples were prepared by varying fiber percentage as 0%, 0.3%, 0.6%

and 0.9% by dry weight of soil for the both types of fibers and the samples were tested

for unconfined compressive strength, volumetric shrinkage, free swell and swelling

pressure. Test results indicated that fiber reinforcement enhanced strength and reduced

volumetric shrinkage and swelling pressure. Fiber reinforcement also decreased swell

potential of the soil.

Investigations were also carried out with nylon fibers and natural fibers having

different aspect ratios (Al-Akhras et al. 2008) to study the influence of fibers on swelling

properties of clayey soils. Four aspects ratios (l/d) of 25, 50, 75 and 100 and five

different fiber contents of 1%, 2%, 3%, 4% and 5% were used in the study. Results

revealed that the swelling pressure and swell potential reduced with increase in the fiber

content significantly. It was also observed that natural fiber is more efficacious than

nylon fibers. Further, a lower aspect ratio appeared to have a greater effect in reducing

swelling pressure in both the types of fibers.

Viswanatham et al. (2009) studied the effect of polypropylene type fibers on

swelling behaviour of expansive soils. One-dimensional swell tests were conducted on

remoulded expansive soils without reinforcement and with reinforcement in the

percentage range of 0.25% and 0.5% and at fiber lengths of 30mm, 60mm and 90mm.

21

22

Digital imaging technique as well as dial gauge was used to measure the heave of

reinforced and un-reinforced expansive clay beds. Reduction in heave was proportional

to fiber content and maximum heave was observed at low aspect ratio at both the fiber

contents of 0.25% and 0.5%. It was also observed that the length of the fiber was the key

factor that influenced the reinforcing effect of fiber. Discrete and randomly distributed

fibers were found efficacious in reducing heave.

Investigation on mixture of fiber and fly ash was also carried out (Puppala, 2001)

to stabilize expansive soils. This technique was also found effective in reducing plasticity

and free swell characteristics. Kumar et al. (2007) studied the influence of fly ash, lime

and polyester fibers on compaction and strength properties of expansive soils. Different

percentages of randomly oriented fibers were introduced in mixes such as soil- lime,

soil-fly ash, and soil-lime-fly ash and compaction tests, unconfined compression tests

and tensile strength tests were conducted on specimens cured for 7, 14 and 28 days. It

was observed that strength increased with increase in curing period. Table 2.10 shows

the summary of test results.

23

Type of fiber

% of fiber Types of tests conducted on expansive soils

Conclusions References

Polypropylene

fiber

0%, 0.3%, 0.6% and

0.9%

0.25% and 0.5%

aspect ratio: 15, 30

and 45

Unconfined compressive

strength, volumetric

shrinkage, free swell and

swelling pressure

1-D swell test

Unconfined compressive strength

increased whereas volumetric shrinkage

and swelling pressure decreased

Reduction in heave was proportional to

fiber content;

Discrete and randomly distributed fibers

were found efficacious in reducing heave

Puppala and

Musenda (2000)

Viswanatham et al.

(2009)

Nylon fiber

1%, 2%, 3%, 4% and

5%

aspect ratio: 25, 50,

Swelling pressure and

swell potential

Swelling pressure and swell potential

decreased with increase in fiber content;

Low aspect ratio has greater effect in

Al-Akhras et al.

(2008)

Table 2.10 Summary of data on fiber reinforcement of expansive soils

75 and 100

0%, 1% and 2%

aspect ratio: 15

Heave test

reducing swelling pressure

Heave decreased as the number of

wetting-drying cycles increased

Heave decreased in both fiber-reinforced

and unreinforced conditions.

Phanikumar et al.

(2011)

Polyester fiber

0.5%, 1.0%, 1.5%

and 2%

Fiber length : 3mm,

6mm and 12mm

Compaction , unconfined

compressive strength,

tensile strength

Split tensile strength and unconfined

compressive strength increased with

increase in fiber content.

Kumar et al. (2007)

24

2.3.3. ADOPTING SPECIAL FOUNDATION TECHNIQUES

Apart from the above mentioned physical alteration and chemical alteration

techniques, there are some special foundation techniques adopted in expansive soils.

They are tension- resistant foundations, devised to absorb the tensile force generated in

expansive clay beds during swelling. Some of the special foundation techniques

employed to counteract heave in expansive soils are explained below:

2.3.3.1 Drilled piers

Piers of small diameter but of considerable length are drilled (Fig 2.2) into

expansive soils up to a zone where there is no effect of moisture change (Chen, 1988).

These are straight-shafted piers with a uniform diameter. Both the end bearing resistance,

E (psf) and friction resistance of drilled piers increase with depth at the rate of about 10%

per meter. Therefore, the load carrying capacity, Q (kips) of a drilled pier is given (Chen,

1988) as,

Q = [ A’ (E+0.03 EL’) + 0.1(E+0.03 EL) C’L’] (2.11)

where A’ = Area of cross-section of the pier (sq.ft.) and

L’ = Depth of penetration into the bearing stratum (ft.) and

C’ = Perimeter of the pier (ft.).

The uplift force, U (lb) is given by,

U = 2πr’fu(D’-d’) (2.12)

where r’ = radius of the pier (ft.),

d’ = depth of the zone of soil unaffected by wetting (ft.),

D’ = total length of the pier (ft.),

U = Swelling pressure (psf) and

fu = coefficient of uplift between concrete and soil

The resisting force comprises of the dead load of the pier and the skin friction and is

given by

25

W= (π r’2p + 2πr’s’d’) (2.13)

where p = Unit dead load pressure (psf),

s’ = skin friction surrounding the pier (psf),

w = total withholding force (lb),

d’ = depth of the zone of soil unaffected by wetting (ft) and

r’ = radius of the pier (ft).

A rational pier formula is obtained by equating the uplift force and the resisting force.

Therefore,

2πr’fu(D’-d’) = (πr’2p+2πr’s’d’)

or, p = 2/r’[fu(D’-d’) –s’d’] (2.14)

Reinforcement for tension Grade beam

Dead load

Air space beneath Grade beam

Uplift pressureD-d Zone of wetting

D

Zone of soil unaffected by wetting

d

Fig. 2.2 Drilled piers

26

2.3.3.2. Belled piers

Belled piers have an enlarged bell-shaped base at the end of a straight pier shaft

for increasing the load carrying capacity. Such piers such as having enlarged diameter at

the bottom are referred to as belled piers (Fig 2.3). The swelling soil tends to grip the

pier shafts and lift them. The uplift force ‘U’ exerted on the belled pier is given as

U = ( P + Fw + Fs) (2.15)

where P = total vertical pressure on the pier,

Fw = total weight of the soil above the bell,

and Fs = total shearing resistance along the circular failure surface.

The most prominent disadvantage of belled piers is the cost and the difficulty of

inspection.

2.3.3.3. Friction piers

Friction piers can be used where the bedrock is very deep and the upper layers of

the soil are expansive (Chen, 1988).

Reinforcement for tensionGrade beam

Dead load

Air space beneath Grade beam

Uplift pressure Zone of wetting (D-d)

D

Zone of soil unaffected by wetting (d)

Assumed circular plane of failure

Fig. 2.3 Belled piers

27

The ult

(2.16)

ned from field load tests.

ould not be encountered within the length of the pier.

India) and designed for Indian conditions. Under-reamed

piles a

ansfer the load of a building

through the piles to a depth beyond the zone of seasonal moisture variation or active

zone.

inactive zone

Fig 2.4 Under-reamed pile foundation

imate shaft resistance (S) for a homogeneous clay stratum is given by

S = πr’2α’Su

where r = radius of the pier,

Su = undrained shear strength, and

Α’ = a reduction factor, obtai

The disadvantage is that soft soil sh

2.3.3.4. Under- reamed piles

Under-reamed piles are similar to that of belled piers. Under-reamed piles were

developed by CBRI, Roorkee (

re most effective in isolating the structure from foundation soil. Fig (2.4) shows

the schematic diagram of an under-reamed pile. They have an enlarged bottom anchored

in the inactive zone as shown. This resists the uplift force on the foundation. Sometimes

double under-reams are also cast.

Grade beam

under-ream with enlarged bottom

active zone of expansive clay bed

Under-reamed piles

The principle involved in under-reamed piles is to tr

Under-reamed piles are usually bored cast in situ piles bored down to at least

0.6m into the inactive zone with the lower portions reamed in the form of a bulb. The

28

spacing of the piles may range from 1.5m to 3.0m. The piles are connected at their top by

plinth beams which support the super structure.

Some of the techniques discussed above have some limitations and some

techniq

.4. GRANULAR PILES

.4.1. INTRODUCTION

ed as a ground improvement technique to improve soft clays

and loo

.4.2. FAILURE MODE

weak soil when loaded vertically on top may fail with zero

lateral

greater than critical length (Madhav et al. 1994).

ues are not economical for lightly loaded structures. Hence, a simple technique in

the form of a granular pile anchor (GPA) was suggested as a new foundation technique

for expansive soils (Phanikumar, 1997; Phanikumar and Rao, 2000; Phanikumar et al.

2004; Rao et al. 2007; Rao et al. 2008) is explained in the following section.

2

2

Granular piles are us

se cohesionless deposits (Thornburn and Mc Vicar, 1968; Greenwood, 1970;

Baumann and Bauer, 1974; Hughes and Withers, 1974; Rathegeb and Kutzner, 1975;

Hughes et al. 1975; Priebe, 1976; Balaam et al. 1977; Aboshi et al. 1979; Goughnour

and Bayuk, 1979; Rao and Bhandari, 1980; Datye and Nagaraju, 1981; Rao, 1982;

Ranjan and Rao, 1983; Rao and Ranjan, 1985; Rao, 1986; Ranjan, 1989). The response

of granular pile reinforced ground under load is assessed in terms of its bearing capacity

and settlement.

2

A granular pile in a

strain (Madhav and Vitkar, 1978), by bulging (Hughes and Withers, 1974; Rao

and Bhandari, 1980; Datye and Nagaraju, 1981; Rao, 1982; Ranjan and Rao, 1983; Rao

and Ranjan, 1988), by general shear failure (Greenwood, 1970; Madhav and Vitkar,

1978) and by sliding (Aboshi et al. 1979). An end- bearing granular pile of length greater

than 3 or 4 diameters fails in bulging (Hughes and Withers, 1974). The length of the

bulge is limited to 4-5 pile diameters. A very short end-bearing granular pile undergoes

either a general or local shear failure at the surface. For the soil conditions encountered

in practice, bulging is the controlling failure mechanism of most granular piles of length

29

Failure in the case of a group of granular piles is by a combination of bulging and

local shear failure. Groups with short column lengths fail in end bearing. Spacing of piles

in a gro

The ultimate bearing capacity of the ground treated with granular piles is

elow:

2.4.3.1

e pressure of which resists these

b

pressure of the granular material.

qual to the depth of the footing from the ground level

up depends on the degree of improvement and settlement tolerance for the design

load (Watt et al. 1967; Engelhardt and Kirsch, 1977). If granular piles are closely spaced

(sp < 5d), settlements are significantly reduced (Balaam et al. 1977) and failure avoided.

Bulging of granular piles can be avoided by replacing the bulged portion of the pile by

concrete plugs or cement grout which is called skirting (Engelhardt and Kirsch, 1977;

Floss, 1979). Provision of all round surcharge is also suggested to increase passive

resistance and consequently the load bearing capacity. Recent research (Alamgir, 1989;

Adayat and Hanna, 1991; Madhav et al. 1994) has shown that granular piles either

reinforced by geogrids or enveloped by a geomembrane perform better as bulging is

prevented and ultimate capacity is increased. The effectiveness of granular piles will be

better if they are combined with other soil-replacement techniques (Bergado et al. 1991).

2.4.3. ULTIMATE BEARING CAPACITY

determined by different approaches explained b

. Passive pressure approach: In this approach, the pile material dilates and exerts

lateral stresses on the surrounding clay, the passiv

stresses (Greenwood, 1970). Hence, the technique of granular piles is feasible only in

soils of undrained strength more than 15 kPa (Barksdale and Bachus, 1983; Juran and

Guermazi, 1988). The ultimate bearing capacity (qult) of a single granular pile is equal to

the ultimate lateral strength of the surrounding soil (Hughes and Withers, 1974).

Therefore (2.17)

pupbpult KcKZPq 2+== γ

where, γ is the bulk density of clay,

Z is the total depth of bulge and

Kp is the coefficient of passive earth

The total depth of bulge (Zb) is e

plus the depth of the bulge of the pile.

30

2.4.3.2. General shear failure approach: Madhav and Vitkar (1978) considered the

plane strain version of a granular pile as a granular trench and proposed the failure

(2.18)

(2.19)

and

Nc1, Nc2 ensionless factors, depending on the properties

ch material. c1 and γ1 are cohesion and density of the trench material, and c2 and

ssible, rigid plastic column contained in a semi-infinite, rigid, plastic

ter test as,

p1 is the lateral limit stress in a pressuremeter.

Unit-cell approach: Priebe (1976) and Goughnour and Bayuk (1979) considered

ar pile is

ribution of

(2.22)

mechanism. The theory is developed for a c-φ soil. The ultimate capacity (qult) is given

as

where

qfcult NDBNNcq 222 21 γγ γ ++=

212

1cCc NN

ccN +=

211

γγγγ NNN +=

2

(2.20)

of soil , Nγ1, Nγ2 and Nq are dim

and tren

γ2, those of soil.

2.4.3.3. Pressuremeter approach: In this approach, the granular pile is considered as a

single, incompre

soft soil (Schlosser and Juran, 1983). The lateral limit stress, σr, is found from triaxial

compression test as,

σr = 2cu +σs (2.21)

or from a pressureme

σr = p1

where σs is the normal stress and

2.4.3.4

a single granular pile and its surrounding tributary soil as a unit cell. The granul

assumed to be rigid, plastic and incompressible and the soft soil to be elastic.

As the vertical settlement of the granular pile and the soil is approximately the same

(Vautrain, 1977), stress concentration occurs in the granular pile. The dist

vertical stress within a unit cell is expressed by a stress concentration factor, n, as

s

p

qq

n =

31

where qp and qs are stresses in the pile and the soil respectively.

For equilibrium, ) at a given depth of the unit cell is given as,

(2.23)

s = (Area of the pile) / (Area of the unit cell)

(2.24)

(2.25)

pproxima itely long cylinder expanding into the soil, and

(2.26)

,

(2.27)

the average stress (q

q = qp as+ qs(1-as)

where ‘as’ is area replacement ratio given as,

a

The stresses in the pile and soil are given as,

s

p anq

)1(1 −+=

qn .

and

ss an

qq)1(1 −+

=

2.4.3.5. Cavity expansion approach: The lateral bulging of the granular pile is

a ted in this approach as an infin

the confining pressure on the pile evaluated (Hughes and Withers, 1974). Based on the

elastic –plastic theory given by Gibson and Anderson (1961) for a frictionless material,

the ultimate lateral stress (σ3) of the surrounding soil is given (Hughes and Withers,

1974) as,

⎥⎤

⎢⎡

++= log1σσ Ec C

where σro is the total in-situ lateral stress,

Ec is the elastic modulus of the soil,

c is the undrained shear strength and

µ is the poisson’s ratio.

The ultimate capacity (qult) is given as

⎦⎣ + )1(203 µCer

( ) ⎟⎟⎠

⎞⎜⎜⎝

⎛−+

⎭⎬⎫

⎩⎨ ⎥

⎦

⎤⎢⎣ +

++=s

sceroult c

cqϕϕ

µσ

sin1sin1

12log1

⎧ ⎡ E

where φs is the angle of internal friction of the pile material.

32

Vesic (1972) extended this theory to soils with both friction and cohesion and gave an

(2.28)

ess (σ1+σ2+σ3) /3 at the equivalent failure depth and

cording to Vesic (1972), is given as

(2.29)

.4.3.6. Expe ithers (1974) conducted a series of

d Withers (1974) is given as,

(2.30)

ere σro is and Kp is the coefficient of passive pressure, which

Bayuk (1979), Rao (1982) and Greenwood and Kirsch (1983)

showed

.4.4. SETTLEMENT OF FOUNDATIONS ON COMPOSITE GROUND

ments of

compo

method (Aboshi et al. 1979; Barksdale, 1981),

[ ]

expression for the ultimate lateral resistance (σ3) of the surrounding soil as,

σ3 = cF’c + qF’q

where ‘c’ is the cohesion

‘q’ is the mean str

F’c and F’q are cavity expansion factors.

The ultimate capacity of stone column, ac

⎟⎟⎞

⎜⎜⎝

⎛−+′+′=

s

sqCult FqFcq

ϕϕ

sin1sin1

⎠

2 rimental approaches: Hughes and W

model tests on single granular piles in normally consolidated clay and found that the

ratio of the applied stress to the undrained shear strength in the case of treated ground

was more than that in the case of untreated clay.

The ultimate pile capacity predicted by Hughes an

uro

ult cq

4=σ

pK+

wh the in-situ radial stress,

varies from 4 to 5.89.

Goughnour and

that the shear stresses at the pile-soil interface would be small as there is no

relative movement of the pile with respect to soil.

2

The analytical and experimental approaches to predict the settle

site ground are explained below:

Analytical approaches: In the equilibrium

the load applied to the unit cell is equally shared by the pile and the soil. The settlement

(st) over a depth ‘H’ of the treated ground is calculated from the conventional equation

from consolidation settlement.

33

The settlement reduction ratio (β´) which is the ratio of the settlements of treated

and untreated grounds ( st and s) is given in terms of average effective stress(σo) and

applied stress (σ) as,

(2.31)

where

(2.32)

Priebe (1976) also considered the unit cell concept and estimated reduction in

settlement. The change in volume within the soil is directly related to vertical shortening

of the pile. Both the pile and soil are assumed to undergo equal settlements.

Goughnour and Bayuk (1979) and Goughnour (1983) assumed the granular pile

to be in plastic equilibrium and the volume changes to be accommodated by the

surrounding clay. The vertical strain are calculated assuming the pile to be linearly

elastic.

Based on unit cell concept, Van Impe and De Beer (1983) predicted ‘β’ considering the

granular piles as

(i) plastic, incompressible columns and

(ii) linearly elastic columns.

2.4.4.1. Experimental approach: Hughes et al. (1975) measured radial strains from the

pressuremeter tests and calculated the vertical strain within the granular pile as twice the

radial strain. The settlement is found by dividing the pile into layers of different

thicknesses.

Floss (1979) added the vertical displacement of the top of the pile to the

settlement of the soil strata below the pile tip.

The use of plain granular piles, at times with a skirting too, was confined to non-

swelling soils only. However, Rao (1986) suggested the use of these piles in expansive

soil deposits also as an alternative to the existing foundation practices, without

supporting it with any laboratory or field investigations. But, a mere granular pile, which

⎪⎭⎪⎩⎟⎠

⎜⎝ 0

10 σ⎪⎬

⎫

⎪⎨

⎧

⎟⎞

⎜⎛ +

⎞⎛ +

==′0

0

log σσ

σµσ

βs

t ⎪⎪

⎪⎪ ⎟⎟

⎠⎜⎜⎝

0010log

σs

sc

1=µ

an )1(1 −+

34

is an un-cemented particulate medium, cannot counteract heave of a foundation caused

by a swelling soil. Hence, Phanikumar (1997) modified a granular pile as a granular-pile

anchor (GPA) by anchoring the foundation at the bottom of the granular pile to an anchor

plate through an anchor rod and rendered it into a tension-resistant foundation system.

Phanikumar (1997) conducted an extensive laboratory testing on small scale GPAs and

found the GPA system extremely efficacious. .

2.5. GRANULAR PILE-ANCHORS (GPA) Phanikumar (1997) suggested a simple and a new foundation technique in the

form of granular pile anchor (GPA), by modifying a granular pile into a tension-resistant

foundation through an anchor plate and an anchor rod.

A granular pile-anchor (GPA) is an innovative modification of the conventional

granular pile. In a GPA the foundation is anchored to an anchor plate (or base plate) at

the bottom of the granular pile through a mild steel anchor rod. Fig 2.5 shows a

schematic representation of a GPA. Because of the anchor introduced in the granular

pile, the system becomes a tension-resistant foundation. Hence, the tensile uplift force

(Pu) caused on the foundation by the swelling soils is resisted by downward friction,

mobilized over the cylindrical pile-soil interface. Fig 2.6 shows the forces acting on a

GPA.

Fig 2.5 Concept of a granular pile-anchor (GPA)

Footing Anchor rod Granular pile-anchor Granular fill Anchor plate

35

The uplift force (PU) is caused by the swelling pressure (ps) of the soil and the

resistance (PR) to uplift is due to

(i) the weight of the GPA (Wgpa) acting in the downward direction and

(ii) the shear resistance or the frictional resistance mobilized over the cylindrical

pile-soil interface due to the shear parameters of the interface, namely, c´ and

φ´

In the case of a single GPA, the uplift force (Pu) on the foundation due to

swelling pressure (ps) of the soil can be written as,

(2.33)

where, Df is the diameter of the foundation and

Dgpa is the diameter of GPA.

However, if there are ‘n’ number of GPAs under a footing, Pu can be written as

(2.34)

Footing Df

Lgpa

Uplift force (Pu) due to swelling pressure on the footing

Granular pile-anchor

Anchor plate

Anchor rod

Fig 2.6 Forces acting on a GPA

Resistance to uplift (PR)

Lateral swelling pressure (Ksps)

⎟⎠⎝ 44 f⎞

⎜⎛ −= 22

gpasu DDpP ππ

⎟⎠

⎜⎝

−= 22

44 f DnDpP ⎞⎛ × gpasuππ

Dgpa

36

where ‘n’ is the number of GPAs in the group anchoring the foundation,

Similarly, in the case of single GPA, the resisting force (PR) can be written as,

(2.35)

( )( )[ ]φσπ ′++′+= tanKsPsLDP KcW vogpagpagpaR

where,

Wgpa is the weight of the GPA

Dgpa is the diameter of GPA,

Lgpa is the length of the GPA,

K is the lateral earth pressure coefficient,

Ks is the lateral swell pressure coefficient and

is the effective overburden pressure. ′

voσ

If there are ‘n’ numbers of GPAs anchoring a footing, the resisting force (PR) can be

written as

(2.36) ( )( )[ ]nKcWR KsPsLDP vogpagpagpa φσπ ′++′+= tan

Hence the factor of safety (FS) against uplift is

(2.37) u

R

PPFS =

A series of laboratory scale tests was conducted on expansive clay beds

reinforced with GPAs of varying diameter (Dgpa), length (Lgpa) and relative density (Dr)

to study the efficacy of this system (Phanikumar, 1997). Heave tests were conducted, and

rate and amount of heave were studied in the light of the above parameters. Test results

showed that the rate of heave improved with the introduction of GPA in the clay beds,

and the amount of heave of the clay beds also decreased significantly (Phanikumar,

1997; Phanikumar et al. 2004).

Rao et al. (2007), Rao et al. (2008) and Phanikumar et al. (2008) conducted field-

scale test program to study the pullout response of GPAs embedded in expansive clay

beds. Pullout tests were conducted for varying lengths and diameters. The length of the

granular pile anchor was varied as 500 mm, 750 mm and 1000 mm. The diameter of the

37

granular pile anchor was varied as 100mm, 150mm and 200mm for each length of the

GPA. Of the various single granular pile anchors, the GPA with l/d ratio of 5 showed the

best pullout load response when tested alone, resulting in a failure uplift capacity of

14.7kN. One pullout test was also performed to study the group effect on the behaviour

of GPAs. In the group test five GPAs were used with a c/c spacing of two times the

diameter of the GPA. All the GPAs were of diameter 150mm and length 1000mm. The

GPA at the center of the group was loaded. The granular pile anchor under the effect of a

group of GPAs resulted in increased resistance to uplift load for a given upward

movement. The failure load of the GPA was 18kN as against a failure pullout load of

12kN for the GPA when tested single indicating a 50% improvement in the failure load.

Phanikumar et al (2008) conducted an extensive field scale study on GPAs and found

that heave was reduced significantly in the field scale GPAs also. Rao et al. (2008)

subjected field scale GPAs to compressive loading and obtained excellent results. It was

found that the bearing capacity of the clay beds reinforced by field scale GPAs improved

significantly.

The investigations carried out on granular pile anchors (GPAs) reveal that the

GPA system is an effective foundation technique for reducing heave in expansive soils.

It is also a simple and cost-effective technique. The behaviour of GPA during swelling

was thoroughly investigated but its behaviour during shrinkage has not been studied so

far. It is important to understand the behaviour of GPA-reinforced expansive clay beds

subjected to swelling and shrinkage also as they are under the influence of alternate

swelling and shrinking due to the changes in climatic conditions. This thesis presents

research results from a detailed experimental investigation conducted to understand the

behaviour of expansive clay beds subjected to swelling and shrinkage cycles. Effect of

number of GPAs in an expansive clay bed (n = 0, 1, 2 and 3) and the number of swell-

shrink cycles (N = 0, 1, 2 and 3) on the amount of heave and shrinkage was studied.

Useful results were obtained.

The details of experimental investigations carried out are presented in Chapter 3

and the analysis of test results in the subsequent chapters.

38