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Transcript of Recommended LRFD Guidelines for the Seismic Design of Highway Bridges
1325 NCHRP 20-7(193) Task 6 Report.doc
NCHRP 20-07/Task 193
Task 6 Report for
Updating
“Recommended LRFD Guidelines for the Seismic Design of Highway Bridges”
Imbsen & Associates, Inc. - A TRC Company
1325 NCHRP 20-7(193) Task 6 Report.doc i
TABLE OF CONTENTS
Section No. Page No.
1 Review Reference Documents ............................................................ 1-1
2 Finalize Seismic Hazard Level............................................................. 2-1
2.1 Recommended Approach to Addressing Seismic Hazard ...................... 2-1
2.1.1 Background on Seismic Hazard................................................... 2-2
2.2 Proposed Seismic Hazard for Design of Normal Bridges ....................... 2-2
3 Expand the Extent of the “No Analysis” Zone.................................. 3-1
3.1 Introduction............................................................................................. 3-1
3.2 Proposed Range of Applicability of Analysis .......................................... 3-3
3.3.1 Column Shear Requirement for SPC B ..................................... 3-12
3.3.2 Column Shear Requirement for SPC C ..................................... 3-14
3.4 Drift Capacity for SPC B and SPC C .................................................... 3-15
3.5 Hinge Seat Requirement ...................................................................... 3-18
3.5.1 Minimum Edge Distance............................................................ 3-18
3.5.2 Other Movement ........................................................................ 3-19
3.5.3 Skew Effect................................................................................ 3-20
3.5.4 Relative Hinge Displacement..................................................... 3-21
1325 NCHRP 20-7(193) Task 6 Report.doc ii
4 Select the Most Appropriate Design Procedure for Steel ................. 4-1
4.1 General................................................................................................... 4-1
4.2 Design Examples.................................................................................... 4-2
4.3 Load Path and Performance Criteria ...................................................... 4-4
4.4 Summary ................................................................................................ 4-8
5 Recommend Liquefaction Design Procedure .................................... 5-1
5.1 Objective ................................................................................................ 5-1
5.2 NCHRP 12-49 Liquefaction Design Requirements................................. 5-1
5.3 Damage Severity in Past Earthquakes ................................................... 5-3
5.4 Proposed Liquefaction Design Requirements ........................................ 5-4
5.5 Summary ................................................................................................ 5-6
1325 NCHRP 20-7(193) Task 6 Report.doc iii
LIST OF FIGURES
Figure No. Page No.
Figure 2-1: Idealized Load – Deflection Curve of a Bridge................................... 2-6
Figure 2-2: Probabilistic to Deterministic Ratio at Selected Sites..................... 2-12
Figure 3-1: Elastic Response Spectra Curves (5% Damping) for Soil Profile Type D (M = 6.5 ± 0.25) (Caltrans SDC)................................... 3-6
Figure 3-2: Elastic Response Spectra Curve (5% Damping) for Soil Profile Type D (M = 8.0 ± 0.25) (Caltrans SDC)................................... 3-7
Figure 3-3: Core Flowchart ................................................................................... 3-11
Figure 3-4: Proposed Drift Capacity for SPC B and C ........................................ 3-18
Figure 3-5: Skew Effect Seat Width Amplification Factor for Various Skew Angles ................................................................................................. 3-20
Figure 3-6: Relative Seismic Displacement vs. Period Ratio ............................. 3-23
Figure 3-7: Proposed Seat Width Compared to NCHRP 12-49 and DIV 1A (H=20ft) ................................................................................................ 3-25
Figure 3-8: Proposed Seat Width Compared to NCHRP 12-49 and DIV 1A (H=30ft) ................................................................................................ 3-25
Figure 4-1: Seismic Load Path and Affected Components .................................. 4-6
1325 NCHRP 20-7(193) Task 6 Report.doc iv
LIST OF TABLES
Table No. Page No.
Table 2-1: Identified Sources of Conservatism ...................................................... 2-4
Table 2-2: Selected Sites for PSHA/DSHA Comparison ........................................ 2-8
Table 2-3: Design Spectral Acceleration based on NCHRP 1997 ......................... 2-9
Table 2-4: Probabilistic Spectral Acceleration for 10% and 5% in 50 Years..... 2-10
Table 2-5: Spectral Acceleration (Type B & D Soil) for 5% in 50 Years.............. 2-10
Table 2-6: One-Second Spectral Acceleration Comparison to USGS 1996 ...... 2-11
Table 2-7: Probabilistic to Deterministic Comparison of One-Second Acceleration........................................................................................ 2-12
Table 3-1: Proposed Partitions for Seismic Performance Categories A, B, C, and D............................................................................................................. 3-4
Table 3-2: 1.0 sec. Spectral Acceleration for Magnitude 6 .................................... 3-8
Table 3-3: 1.0 sec. Spectral Acceleration for Magnitude 7 .................................... 3-8
Table 3-4: 1.0 sec. Spectral Acceleration for Magnitude 8 .................................... 3-8
Table 3-5: 1.0 sec. Spectral Acceleration (Division 1A)......................................... 3-8
Table 3-6: Seismic Performance Category for Selected Sites .............................. 3-9
Table 3-7: Column Parameters .............................................................................. 3-17
1325 NCHRP 20-7(193) Task 6 Report.doc v
Table 4-1: Reduction Factors for Steel Superstructure Bracings......................... 4-8
Table 5-1: Damage Severity Description................................................................. 5-3
Table 5-2: Damage Severity Rating vs. Earthquake Magnitude............................ 5-4
1325 NCHRP 20-7(193) Task 6 Report.doc 1-1
TASK 1
1 REVIEW REFERENCE DOCUMENTS
A review of the pertinent documents and information that were available was
conducted and has been included in Tasks 2 thru 5 as needed. The reference
material that was selected for inclusion is attached as appendices for each of
the individual tasks. Their inclusion as appendices makes this Letter Report
somewhat self-contained and additionally, makes it more convenient for our
reviewers.
A separate section is included in this Letter Report for each of the tasks as
described below:
Section 2 presents the justification for the 1000-year return period (i.e., 5%
probability of exceedance in 50 years) as recommended for the seismic design of
highway bridges.
Section 3 includes a description of how the “no analysis” zone is expanded and
how this expansion is incorporated into the displacement based approach.
Section 4 describes the two alternative approaches available for the design of
highway bridges with steel superstructures and concludes with a
recommendation to use a force base approach for the proposed specification.
Section 5 describes the recommended procedure for liquefaction design to be
used for highway bridges. This aspect of the design is influenced by the
recommended hazard level and the no analysis zone covered in Tasks 2 and 3
1325 NCHRP 20-7(193) Task 6 Report.doc 1-2
respectively. The recommendations proposed are made taking into account the
outcome of these two tasks for Seismic Performance Category D.
1325 NCHRP 20-7(193) Task 6 Report.doc 2-1
TASK 2
2 FINALIZE SEISMIC HAZARD LEVEL
2.1 Recommended Approach to Addressing Seismic Hazard
The recommended approach to addressing the seismic hazard is based on the
following positions:
• Recommendations would be Primarily for Design against the Effects
Ground Shaking Hazard
• Selection of a Return Period for Design less than 2500 Years
• Inclusion of the USGS 2002 Update of the National Seismic Hazard
Maps
• Effects of Near Field and Fault Rupture to be addressed in a separate
later Task.
• Displacement Based Approach with both Design Spectral Acceleration
and corresponding Displacement Spectra provided
• Hazard Map under the control of AASHTO with each State having the
option to Modify or Update their own State Hazard using the most
recent Seismological Studies consistent with the Established Risk
1325 NCHRP 20-7(193) Task 6 Report.doc 2-2
2.1.1 Background on Seismic Hazard The current State of the Practice in addressing the seismic hazard for the
design of bridges in the U.S. has evolved from just conforming to AASHTO
Division 1-A requirements to adopting higher standards that take into account
the possible effects of larger earthquakes in the Eastern United States and the
impacts of major earthquakes that occurred recently in the Western United
States, Japan, Taiwan and Turkey. This change in the Seismic Hazard
Practice can be best illustrated in looking at the following sources:
• NEHRP 1997 Seismic Hazard Practice
• Caltrans Seismic Hazard Practice
• NYCDOT and NYSDOT Seismic Hazard Practice
• NCHRP 12-49 Seismic Hazard Practice
• SCDOT Seismic Hazard Practice
• Site Specific Hazard Analyses Conducted for Critical Bridges
Appendix 2A contains background on seismic hazard drawn from the above
mentioned sources.
2.2 Proposed Seismic Hazard for Design of Normal Bridges
In reviewing the seismic hazard practice in different regions as described
previously, it is apparent that some important aspects of this Practice need to
be taken into consideration when developing new Guidelines. These aspects
are pivotal in reaching the objective of producing Guidelines that are adoptable
by AASHTO.
1325 NCHRP 20-7(193) Task 6 Report.doc 2-3
These aspects include:
1) Consideration for lower return period for Design based on the Maximum
Considered Earthquake (MCE) maps published in 1996 with USGS 2002
Update shall be considered a minimum standard. Modification or
increase in the hazard intensity based on Seismological Studies needs to
be included as an option for states and agencies seeking a higher degree
of hazard identification to a specific region or bridge.
2) The reduction in the design intensity can be implicitly achieved by
considering applying a reduction/modifier factor for design spectrum
derived from USGS MCE maps. An alternative to this approach would
be embarking on developing new maps based on a modified new
definition of the MCE for Bridge Design.
3) Consideration of applying a reduction factor on the hazard intensity for
existing bridges or bridges located in rural areas.
4) Selection of a lower return period for Design is made such that Collapse
Prevention is not compromised when considering historical large
earthquakes. This reduction can be achieved by taking advantage of
sources of conservatism not explicitly taken into account in current
design procedures. These sources of conservatism are becoming obvious
based on recent findings from both observations of earthquake damage
and experimental data. Some of these sources are shown in Table 2-1.
1325 NCHRP 20-7(193) Task 6 Report.doc 2-4
Table 2-1: Identified Sources of Conservatism
Source of Conservatism Safety Factor
Computational vs. Experimental Displacement Capacity of Components
1.3
Effective Damping 1.2 to 1.5 Dynamic Effect (i.e., strain rate effect) 1.2 Pushover Techniques Governed by First Plastic Hinge to Reach Ultimate Capacity
1.2 to 1.5
Out of Phase Displacement at Hinge Seat Addressed in Task 3
1 The conservatism is directly coupled to the seismic reliability of the
structural system under consideration. The current state of the practice
favors continuous superstructures for the majority of bridges with an
objective of minimizing expansion joints to gain functionality, reduce
maintenance, and increase life cycle of the bridge. This selection has a
favorable impact on the earthquake redundancy of the bridge system.
Considering a single performance level of “No Collapse”, the seismic
redundancy of the bridge system is enhanced with the increase of the
number of plastic hinges that must yield and then fail in order to produce
the impending collapse of the structure. This enhanced redundancy
translates into a delayed failure (i.e. collapse) provided sufficient seat
width exists in the bridge system. Therefore two distinctly different
aspects of the design process need to be provided:
a) An appropriate method to design adequate seat width(s) considering out
of phase motions.
b) An appropriate method to design the ductile substructure components
without undue conservatism.
1325 NCHRP 20-7(193) Task 6 Report.doc 2-5
These two aspects are embedded with different levels of conservatism that
need to be calibrated against the single level of hazard considered in the design
process.
The first aspect is highly influenced by variation in the periods of the frames
on both sides of a joint as well as the damping generated by the ductile
behavior of plastic hinges. This aspect is addressed in terms of
recommendations or limits on periods ratio for frames on both sides of an
expansion joint.
The second aspect is addressed using a static push-over analysis. As shown in
Figure 2-4, the collapse displacement is usually reached when the P-Δ line
intersects the load-displacement curve of the structure, because at this point,
any increment in displacement produces an increment in the P-Δ effect due to
gravity loads that cannot be resisted by the lateral resistant system. It is
important to mention for structures with relatively small gravity loads, a much
larger reduction in component strength can be tolerated without reaching
structural collapse. This is especially relevant to bridge columns carrying axial
loads typically ranging from .05 c gf A′ to .15 c gf A′ maximum. In essence, the
continuity of the superstructure and low axial loads in columns make a typical
bridge more resilient against collapse in a seismic event.
1325 NCHRP 20-7(193) Task 6 Report.doc 2-6
Figure 2-1: Idealized Load – Deflection Curve of a Bridge
Under earthquake ground motions at the supports, the structure or any of its
components can fail under a smaller displacement than the displacement
Δcollapse illustrated in Figure 2-1. This failure is mainly attributed to
nonsymmetric cumulative plastic displacement that is highly depended on the
characteristics of the earthquake ground motions. The reliable displacement
capacity is typically associated with the displacement corresponding to a
limited decrease in strength of 20% to 30% maximum obtained under
monotonically increasing deformation. As shown in Figure 2-1, the
displacement capacity Δcapacity can only be established given the descending
slope following the point of maximum lateral resistance Fmax. Recognizing the
1325 NCHRP 20-7(193) Task 6 Report.doc 2-7
complexity of determining Δcapacity, the Δcapacity/bridge is used as a conservative and
simple measure assuming nominal properties.
In summary, the two aspects described above should be considered in the
practice to justify a reduction in the design hazard and ensure the development
of a simplified methodology that addresses the different sources of
conservatism included in the current state of the practice.
In order to assess the feasibility of a reduction in hazard from the 2% in 50
years hazard level adopted by NCHRP 12-49, a Probabilistic/Deterministic
comparison is conducted on 20 sites. Table 2-2 shows the state, city, dominant
source, latitude and longitude of the selected sites. Table 2-3 shows the short
period and one-second spectral acceleration for the Deterministic Seismic
Hazard Analysis (DSHA), the Deterministic Cap taken at 1.5 times the
(DSHA) value, the Maximum Considered Earthquake, and the Design Spectral
Acceleration SDS and SD1 based on NEHRP 1997 guidelines.
Table 2-4 shows the short period and one-second acceleration based on a
Probabilistic Seismic Hazard Analysis (PSHA) for 10% and 5% exceedance in
50 years. Table 2-4 includes two additional sites to the 20 sites identified in
Table 2-2 and 2-3.
Table 2-5 shows the short period and one-second period acceleration including
Type D soil effect for the proposed 5% exceedance in 50 years Design Spectrum.
Table 2-6 shows a comparison of the one-second acceleration (PSHA) to the
USGS 1996. As seen from Table 2-6, California sites show a decrease of the
acceleration values while other sites show a marginal change or an increase.
Table 2-7 shows the PSHA/DSHA comparison for the one-second acceleration of
the PSHA/DSHA ratio at each of the selected sites. These ratios are shown
graphically in Figure 2-2.
1325 NCHRP 20-7(193) Task 6 Report.doc 2-8
Based on this comparison, the following recommendations are proposed:
1. Adopt the 5% in 50 years hazard level for development of a design
spectrum.
2. Ensure sufficient conservatism (1.5 safety factor) for minimum seat
width requirement. This conservatism is needed to enable to use the
reserve capacity of hinging mechanism of the bridge system. This
conservatism shall be embedded in the specifications to address
unseating vulnerability. It is recommended to embed this safety factor
for sites outside of California.
3. Partition Seismic Performance Categories (SPCs) into four categories
and proceed with the development of analytical bounds using the 5% in
50 years Hazard level.
Table 2-2: Selected Sites for PSHA/DSHA Comparison ST CITY FEATURE DOMINANT SOURCE LATITUDE LONGITUDE CA Daly City Zip Code 94015 San Andreas 37.681240 -122.479000 CA San Francisco City Hall San Andreas 37.779083 -122.417450 CA SFOBB Site from Po/Roy Hayward 37.750000 -122.250000 CA Berkeley Site from Po/Roy Hayward 37.871667 -122.271667
1325 NCHRP 20-7(193) Task 6 Report.doc 2-9
CA Benicia Martinez
Site from Po/Roy Concord 38.000000 -122.116667
CA Los Angeles City Hall Puente Hills blind thrust
34.053700 -118.243183
CA Vincent Thomas Site from Po/Roy Location corrected
Palos Verdes 33.749218 -118.271466
CA Long Beach Zip Code 90810 Newport-Inglewood 33.813890 -118.217000 CA Coronado
Bridge Site from Po/Roy Rose Canyon 32.616667 -117.116667
WA Seattle Space Needle Seattle fault zone 47.621150 -122.348950 WA Tacoma North Site from Po/Roy Seattle fault zone 47.250000 -122.366667 UT Salt Lake City State Capital Wasatch fault, Salt
Lake City section 40.776367 -111.887983
UT Salt Lake City Site from Po/Roy Wasatch fault, Salt Lake City section
40.750000 -111.883333
IN Evansville Zip code 47720 New Madrid fault zone 38.023280 -87.617100 MO St. Louis Zip code 63129 New Madrid fault zone 38.466780 -90.319400 KY Paducah Zip code 42003 New Madrid fault zone 37.034190 -88.603800 TN Union City Zip code 38261 New Madrid fault zone 36.428110 -89.059500 TN Memphis City Hall New Madrid fault zone 35.148750 -90.054700 TN Memphis Zip code 38127 New Madrid fault zone 35.225170 -90.008400
Table 2-3: Design Spectral Acceleration based on NCHRP 1997
1325 NCHRP 20-7(193) Task 6 Report.doc 2-10
CITY Det Ss, g Det S1, g 1.5*Det Ss, g 1.5*Det S1, g MCE Ss, g MCE S1, g SDs, g SD1, gDaly City 1.49 0.85 2.23 1.28 2.23 1.28 1.49 0.85San Francisco 0.88 0.44 1.32 0.67 1.50 0.67 1.00 0.44SFOBB 0.84 0.30 1.26 0.45 1.50 0.60 1.00 0.40Berkeley 1.28 0.49 1.93 0.74 1.93 0.74 1.28 0.49Benicia Martinez 0.96 0.31 1.44 0.47 1.50 0.60 1.00 0.40Los Angeles 1.50 0.57 2.24 0.86 2.19 0.74 1.46 0.49Vincent Thomas 1.41 0.63 2.12 0.95 2.08 0.92 1.38 0.61Long Beach 1.28 0.51 1.91 0.77 1.81 0.70 1.20 0.47Coronado Bridge 1.19 0.47 1.78 0.70 1.37 0.54 0.91 0.36Seattle 1.34 0.48 2.01 0.73 1.41 0.48 0.94 0.32Tacoma North 0.47 0.18 0.71 0.28 1.20 0.41 0.80 0.27Salt Lake City 1.28 0.53 1.92 0.80 1.71 0.69 1.14 0.46Salt Lake City 1.25 0.53 1.88 0.79 1.70 0.69 1.13 0.46Evansville 0.27 0.09 0.41 0.13 0.67 0.19 0.45 0.13St. Louis 0.23 0.08 0.34 0.12 0.61 0.17 0.40 0.12Paducah 0.89 0.24 1.33 0.36 1.50 0.47 1.00 0.31Union City 0.86 0.23 1.29 0.35 1.50 0.57 1.00 0.38Memphis 0.60 0.17 0.91 0.25 1.40 0.38 0.93 0.25Memphis 0.65 0.18 0.98 0.27 1.50 0.41 1.00 0.27
Table 2-4: Probabilistic Spectral Acceleration for 10% and 5% in 50 Years
CITY10%/50 yr
Ss, g10%/50 yr
S1, g5%/50 yr
Ss, g5%/50 yr
S1, gDaly City 1.60 0.78 2.15 1.12San Francisco 1.15 0.53 1.45 0.69SFOBB 1.26 0.50 1.57 0.62Berkeley 1.65 0.63 2.19 0.83Benicia Martinez 1.24 0.43 1.58 0.55Los Angeles 1.20 0.41 1.60 0.54Vincent Thomas 1.02 0.37 1.47 0.56Long Beach 0.96 0.35 1.30 0.49Coronado Bridge 0.60 0.22 0.89 0.34Seattle 0.73 0.24 0.99 0.33Tacoma North 0.68 0.23 0.89 0.30Salt Lake City 0.69 0.24 1.10 0.42Salt Lake City 0.68 0.24 1.09 0.42Evansville 0.25 0.07 0.40 0.11St. Louis 0.23 0.06 0.36 0.10Paducah 0.54 0.11 0.97 0.24Union City 0.53 0.12 1.06 0.27Memphis 0.38 0.09 0.75 0.19Memphis 0.40 0.09 0.80 0.20Charleston 0.31 0.06 0.69 0.15Phoenix 0.09 0.03 0.12 0.04
Table 2-5: Spectral Acceleration (Type B & D Soil) for 5% in 50 Years
1325 NCHRP 20-7(193) Task 6 Report.doc 2-11
CITY
Type B5%/50 yr
Ss, g
Type B5%/50 yr
S1, g
Type D5%/50 yr
, g
Type D5%, 50 yr
,gDaly City 2.15 1.12 2.15 1.67San Francisco 1.45 0.69 1.45 1.04SFOBB 1.57 0.62 1.57 0.93Berkeley 2.19 0.83 2.19 1.24Benicia Martinez 1.58 0.55 1.58 0.83Los Angeles 1.60 0.54 1.60 0.81Vincent Thomas 1.47 0.56 1.47 0.84Long Beach 1.30 0.49 1.30 0.73Coronado Bridge 0.89 0.34 1.02 0.58Seattle 0.99 0.33 1.10 0.58Tacoma North 0.89 0.30 1.01 0.54Salt Lake City 1.10 0.42 1.17 0.67Salt Lake City 1.09 0.42 1.15 0.66Evansville 0.40 0.11 0.60 0.26St. Louis 0.36 0.10 0.55 0.24Paducah 0.97 0.24 1.08 0.46Union City 1.06 0.27 1.15 0.50Memphis 0.75 0.19 0.89 0.39Memphis 0.80 0.20 0.94 0.41Charleston 0.69 0.15 0.86 0.34Phoenix 0.12 0.04 0.19 0.09
DSS1DS
Table 2-6: One-Second Spectral Acceleration Comparison to USGS 1996
CITY10%/50 yr
S1, g5%/50 yr
S1, g
10%/50yr1996S1,g
5%/50yr1996S1,g Ratio 1 Ratio 2
Daly City 0.78 1.12 1.08 1.50 1.38 1.35San Francisco 0.53 0.69 0.64 0.83 1.22 1.20SFOBB 0.50 0.62 0.62 0.79 1.24 1.27Berkeley 0.63 0.83 0.65 0.86 1.03 1.04Benicia Martinez 0.43 0.55 0.55 0.69 1.27 1.25Los Angeles 0.41 0.54 0.42 0.54 1.02 1.00Vincent Thomas 0.37 0.56 0.41 0.58 1.10 1.03Long Beach 0.35 0.49 0.42 0.60 1.20 1.23Coronado Bridge 0.22 0.34 0.21 0.31 0.95 0.92Seattle 0.24 0.33 0.22 0.32 0.92 0.97Tacoma North 0.23 0.30 0.20 0.28 0.88 0.93Salt Lake City 0.24 0.42 0.21 0.42 0.86 1.00Salt Lake City 0.24 0.42 0.21 0.40 0.87 0.96Evansville 0.07 0.11 0.06 0.12 0.92 1.08St. Louis 0.06 0.10 0.05 0.10 0.85 0.99Paducah 0.11 0.24 0.09 0.20 0.81 0.83Union City 0.12 0.27 0.09 0.21 0.78 0.79Memphis 0.09 0.19 0.07 0.16 0.81 0.84Memphis 0.09 0.20 0.07 0.17 0.78 0.84Charleston 0.06 0.15 0.07 0.17 1.17 1.11Phoenix 0.03 0.04 0.03 0.04 1.11 1.02
1325 NCHRP 20-7(193) Task 6 Report.doc 2-12
Table 2-7: Probabilistic to Deterministic Comparison of One-Second Acceleration
CITYDet
S1, g5%/50 yr
S1, g RatioDaly City 0.85 1.12 1.31San Francisco 0.44 0.69 1.56SFOBB 0.30 0.62 2.07Berkeley 0.49 0.83 1.67Benicia Martinez 0.31 0.55 1.76Los Angeles 0.57 0.54 0.94Vincent Thomas 0.63 0.56 0.89Long Beach 0.51 0.49 0.95Coronado Bridge 0.47 0.34 0.72Seattle 0.48 0.33 0.68Tacoma North 0.18 0.30 1.63Salt Lake City 0.53 0.42 0.79Salt Lake City 0.53 0.42 0.79Evansville 0.09 0.11 1.24St. Louis 0.08 0.10 1.28Paducah 0.24 0.24 1.00Union City 0.23 0.27 1.13Memphis 0.17 0.19 1.13Memphis 0.18 0.20 1.12
0.00
0.50
1.00
1.50
2.00
2.50
Daly C
ity
San Fran
cisco
SFOBB
Berkele
y
Benici
a Mart
inez
Los A
ngele
s
Vincen
t Tho
mas
Long
Bea
ch
Corona
do B
ridge
Seattle
Tacom
a Nort
h
Salt La
ke C
ity
Salt La
ke C
ity
Evans
ville
St. Lou
is
Paduc
ah
Union C
ity
Memph
is
Memph
is
5%-50 Yr/Deterministic
Figure 2-2: Probabilistic to Deterministic Ratio at Selected Sites
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-1
TASK 2
APPENDIX 2A
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-2
NEHRP 1997 Seismic Hazard Practice FEMA 274 describes the background of NEHRP 1997 provisions as follows:
“The NEHRP Recommended Provisions (BSSC, 1997) update process to the
1994 NEHRP Provisions included the formation of a special Seismic Design
Procedures Group (SDPG), consisting of earth scientists from the USGS and
engineers engaged in the update process. The SDPG was charged with the
responsibility of working with the USGS to produce ground motion maps
incorporating the latest earth science procedures, and with appropriate design
procedures to allow use of these maps in the Recommended Provisions. The
SDPG determined that rather than designing for a nationwide uniform hazard
- such as a 10%/50 year or 2%/50 year hazard- it made more sense to design for
a uniform margin of failure against a somewhat arbitrarily selected maximum
earthquake level”.
“This maximum earthquake level was termed a Maximum Considered
earthquake (MCE) in recognition of the fact that this was not the most severe
earthquake hazard level that could ever affect a site, but it was the most
severe level that it was deemed practical to consider for design purposes. The
SDPG decided to adopt a 2%/50 year exceedance level definition for the MCE in
most regions of the nation, as it was felt that this would capture recurrence of
all of the large-magnitude earthquakes that had occurred in historic times”.
“There was concern, however, that the levels of ground shaking derived for this
exceedance level were not appropriate in zones near major active faults. There
were several reasons for this. First, the predicted ground motions in these
regions were much larger than those that had commonly been recorded by near
field instrumentation in recent magnitude 6 or 7 California events. Second, it
was noted, based on the observed performance of buildings in these
earthquakes, that structures designed ot the code had substantial margin
against collapse for ground shaking that is much larger than that for which the
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-3
building had nominally been designed; in the judgment of the SDPG members,
this margin represented a factor of at least 1.5”.
“Consequently, it was decided to adopt a definition of the MCE in zones near
major active faults that consisted of the smaller of the probabilistically
estimated 2%/50 year motion or 150% of the mean ground motion calculated for
a deterministic characteristic earthquake on these major active faults, and to
design all buildings, regardless of location, to provide for protection of occupant
life safety at earthquake ground shaking levels that are 1/1.5 times (2/3) of the
MCE ground motion”.
Following the 1997 NEHRP Provisions, the ratio of the mapped acceleration at
one-second period for return periods of 474, 1000, 1500, 2000 and 2500 years is
normalized against the mapped acceleration at one-second period for a return
period of 474 years. The results of this normalization for California, Pacific,
Intermountain, Central US, and Eastern US are found in Table 2-1. The
California and Pacific Regions are designated with a deterministic cap based
on the description mentioned in the above paragraphs. The normalization is
appropriate for sites where the short period mapped acceleration SS is greater
than 1.5 g (i.e. higher ground shaking).
Table 2-1: Normalized One Second Spectral Acceleration Return Period
Years Region
California Pacific Intermountain Central US
Eastern US
474 1 1 1 1 1 1000 1.2 1.6 1.6 2.3 2.2 1500 1.4 2.2 2.0 3.5 3.4 2000 1.5 2.6 2.4 4.8 4.5 2500 1.6 3.0 2.7 6.1 5.7
Deterministic CAP
Yes Yes No No No
Achieving a national uniform hazard is difficult given the drastic difference
from one region to the other as illustrated in the normalization shown in Table
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-4
2-1. Furthermore, the regional difference in the recurrence of large magnitude
earthquakes makes the task of achieving a uniform hazard even more difficult.
Therefore, it is important that the selection of the Design Hazard can be
implicitly made such that collapse prevention is not compromised when
considering historical large earthquakes.
Caltrans Seismic Hazard Practice California Practice is described by Caltrans Commentary as follows:
“Caltrans bridge engineering practice has generally embraced deterministic
ground motion hazards since the 1971 San Fernando earthquake.
Deterministic practice considers the largest expected earthquake.
Deterministic practice considers the largest expected earthquake on any
known fault. Caltrans uses the mean event for standard practice, and refers to
it as the maximum credible earthquake (MCE). The deterministic method does
not take into consideration the recurrence of an MCE. This method assumes
that the MCE could occur at any time. Bridge engineering practice, therefore
should prudently allow for structures to be able to resist the MCE demands
without endangering the traveling public.”
“In recent years, Caltrans has implemented alternative ground motion hazard
site evaluations to address special situations. The alternatives were used with
consideration for: a) remaining life of a particular structure; b) bridge
performance capacity following an earthquake; c) liquefaction potential at an
existing bridge site; d) potential hazard at a bridge in anticipation of a future
retrofit or replacement; and other similar situations.”
Caltrans, with the support of an external Seismic Advisory Board and the
ATC-32 project team, has developed a set of seismic performance criteria for
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-5
new bridges. Following ATC-32, all bridges shall be designed to meet the
seismic performance criteria given in Table 2-2. Definitions of the terms in
Table 2-2 are given on the following page.
Table 2-2: Seismic Performance Criteria
Ground Motion
at Site
Ordinary Bridges Important Bridges
Functional-Evaluation
Ground Motion
Service Level-Immediate
Repairable Damage
Service Level-Immediate
Minimal Damage
Safety-Evaluation
Ground Motion
Service Level-Limited
Significant Damage
Service Level-Immediate
Repairable Damage
Each bridge shall be classified as either Important or Ordinary, as follows:
a) Important Bridge: Any bridge satisfying one or more of the following:
– Required to provide secondary life safety
– Time for restoration of functionality after closure would create a
major economic impact
– Formally designated by a local emergency plan as critical
b) Ordinary Bridge: Any bridge not classified as an Important Bridge.
The Evaluation Levels are defined as follows:
a) Safety-Evaluation Ground Motion: This ground motion may be assessed
either deterministically or probabilistically. The deterministic
assessment corresponds to the maximum credible earthquake (MCE), as
defined by the Division of Mines and Geology Open File Report 92-1
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-6
(CDMG, 1992). A probabilistically assessed ground motion is one with a
long return period (approximately 1000-2000 years).
For Important Bridges both methods shall be given consideration; however, the
probabilistic evaluation shall be reviewed by a Caltrans-approved consensus
group. For Ordinary Bridges, the motions shall be based only on the
deterministic evaluation.
b) Functional-Evaluation Ground Motion: This is a probabilistically
assessed ground motion that has a 60 percent probability of not being
exceeded during the useful life of the bridge. The determination of this
event is to be reviewed by a Caltrans-approved consensus group.
The following performance levels, expressed in terms of service levels and
damage levels are defined as follows:
a) Service Levels
– Immediate: Full access to normal traffic is available almost
immediately following the earthquake.
– Limited: Limited access (i.e., reduced lanes, light emergency
traffic) is possible within days of the earthquake. Full service is
restorable within months.
b) Damage Levels
– Minimal Damage: Essentially elastic performance.
– Repairable Damage: Damage that can be repaired with a
minimum risk of losing functionality.
– Significant Damage: A minimum risk of collapse, but damage
that would require closure to repair.
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Following recommendations of ATC-32, Caltrans published the Seismic Design
Criteria Version 1.1, July 1999 (SDC 1.1) with focus on Ordinary Bridges as
previously defined in ATC-32. The design spectra (i.e. ARS curves) included in
SDC 1.1 were adopted from the ATC-32 design spectra. However, recognizing
some of the complexities dealing with the roles of probabilistic and
deterministic assessments, it was found that depending on the seismic activity
of a given region, the deterministic and probabilistic assessments may be
different. In the highly seismic zones of the San Francisco Bay region, the
deterministic ground motion assessments using the mean ARS spectra for the
MCE correspond to return periods of about 300 to 400 years. This variation
between the probabilistic and deterministic approaches is still an outstanding
issue in the California Seismic Hazard Practice.
Based on the currently adopted SDC 1.2 released in December 2001, an
ordinary bridge is designed for a standard 5% damped SDC ARS curve, a
modified SDC ARS curve, or a site-specific ARS curve.
For preliminary design, prior to receiving the geotechnical engineer’s
recommendation, a standard SDC ARS curve may be used in conjunction with
the peak rock acceleration from the 1996 Caltrans Seismic Hazard Map. The
standard SDC ARS curves were adopted from ATC-32. If standard SDC ARS
curves are used during preliminary design, they should be adjusted for long
period bridges and bridges in close proximity to a fault as described below.
For preliminary design of structures within 10 miles (15 km) of an active fault,
the modified SDC ARS curve is obtained by magnifying the spectral
acceleration on the SDC ARS curves as follows:
• Spectral acceleration magnification is not required for T ≤ 0.5 seconds
• Increase the spectral accelerations for T ≥ 1.0 seconds by 20%
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-8
• Spectral accelerations for 0.5 ≤ T ≤ 1.0 shall be determined by linear
interpolation
For preliminary design of structures with a fundamental period of vibration
T ≥ 1.5 seconds on deep soil sites (depth of alluvium ≥ 250 feet (75 m)) the
modified SDC ARS curve is obtained by magnifying the spectral ordinates of
the standard ARS curve as follows:
• Spectral acceleration magnification is not required for T ≤ 0.5 seconds
• Increase the spectral accelerations for T ≥ 1.5 seconds by 20%
• Spectral accelerations for 0.5 ≤ T ≤ 1.5 shall be determined by linear
interpolation
A site specific response spectrum is typically required when a bridge is located
in the vicinity of a major fault or located on soft or liquefiable soils and the
estimated earthquake moment magnitude Mm > 6.5.
In formally adopting the displacement approach following the release of
Caltrans Seismic Design Criteria Version 1.1, July 1999, the State of Practice
is implicitly calibrated to the Mean Hazard as stated above including a 1.2
magnification factor of spectral acceleration ordinates for a period of one
second or greater for bridges near a fault. It is deemed important to mention
that the normalization shown in Table 1 reveals that California’s one-second
spectral ratio is at 1.6 and that the NEHRP 1997 deterministic cap is set at 1.5
of the mean ground motion as previously mentioned under the NEHRP 1997
Seismic Hazard Practice. When considering locations close to active faults, the
normalization of the one-second spectral ratio shows average value is close to
1.8 but can be as high as 2.2 using the USGS 1997 maps. This value
represents an increase of 12% to 38% in comparison to the 1.6 ratio. This
increase is comparable to the 20% increase that the Caltrans Seismic Design
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-9
SDC 1.2 has adopted in its December 2001 release. Recent studies suggest
considerably larger increases that have, however, not been yet endorsed in the
practice
NYCDOT and NYSDOT Seismic Hazard Practice New York State Department of Transportation (NYSDOT) practice continues
in accordance with 1996 AASHTO Standard Specifications for Highway
Bridges, Division 1-A. New York City Department of Transportation
(NYCDOT) has adopted modifications to the 1996 AASHTO Division 1-A that
reflect the findings of the “New York City Seismic Hazard and It’s Engineering
Application” final report prepared by Weidlinger Associates, December 1998.
These modifications are applicable to NYC Metro Region including the
Counties listed in Table 2-3. NYCDOT bridges are classified as Critical,
Essential and Other. Table 2-3 summarizes the relationship of bridge
importance and performance requirements. In all cases, “No Collapse” is
permitted.
The following guidelines are adopted for NYCDOT bridges:
• For Bridges designed by the one level approach (Essential and Other),
Figure 2-1 shows the acceleration response spectra to be used for
different soil types (soil classes). Soil classes are defined in Table 2-4.
• Site specific soil effects for the two earthquake levels approach (i.e.
Critical Bridges) should be obtained from an expert. Soil spectra for
different types of soils, base on NEHRP amplification factors are not
recommended for design of Critical Bridges.
• For Critical bridges, site-specific ground motions shall be computed
using rock motions based on the spectra for hard rock (Soil Class A).
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• For Essential bridges where the site condition can be classified as A, B,
C, D, or E, the empirically derived soil spectra shown on Figure 2-1 (2/3
(2% probability of exceedance in 50 years)) shall be used.
• For bridges classified as Other, including single-span bridges, the
spectra shown on Figure 2-1 shall be used for Soil Classes A, B, C, D,
and E.
• For Soil Class F, regardless of bridge Importance Category, site-specific
analysis should be performed.
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Table 2-3: Performance Criteria and Seismic Hazard Level for Design and Evaluation of Bridges
(Applicable to NYC Metro Region/Downstate Counties: Bronx, Kings, Nassau, New York, Queens, Richmond, Rockland and Westchester)
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-12
Table 2-4: Soil Classification
Figure 2-1: NYCDOT Soil Acceleration Response Spectra for One-
Level Approach (2% in 50 Years Probability of Exceedance) /1.5
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-13
A comparison between the panel spectra adopted by NYCDOT and NEHRP
1997 shows the following:
• The NEHRP/97 MCE hard rock is lower than the Panel Hard Rock as
demonstrated in Figure 2-2.
• The short period soil amplification Panel Spectra factors are lower than
the NEHRP 1997 corresponding factors as demonstrated in
Table 2-5.
• The long period soil amplification Panel Spectra factors are essentially
the same in comparison to the NEHRP 1997 corresponding factors as
demonstrated in Table 2-6.
• In Summary, the long period spectral acceleration values for
NEHRP/1997 are lower than the NYCDOT adopted Panel Spectral
values. However, the difference is less pronounced when comparing the
NEHRP 1997 values to the 2/3 NYCDOT values adopted for essential
and other bridges as mentioned earlier in this section.
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Figure 2-2: Spectra Comparison – NYC Rock Acceleration Response Spectra – 5% Damping-2500 Year Return Period = 2% in 50 Years
Probability of Exceedance
Table 2-5: Fa = Short Periods Amplification Factor/Normalized For Soil Class B For 2% In 50 Yrs Probability of Exceedance Curves (*)
For NEHRP 94 For NEHRP 97 Panel Based Spectra Soil Class Aa = .16g Ss = .40g Aa = .30g Ss = .72g - .75g
A 0.8 0.8
B 1.0 1.0
C 1.2 1.1
D 1.48 1.2
E 2.02 1.2
(*) VALUES FOR Aa AND Ss ARE FOR SOIL CLASS B
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Table 2-6: Fv = Long Periods Amplification Factor/Normalized For Soil Class B For 2% In 50 Yrs Probability of Exceedance Curves (*)
For NEHRP 94 For NEHRP 97 Panel Based Spectra Soil Class Av = .09g S1 = .09g Av = .13g S1 = .13g
A 0.8 0.8
B 1.0 1.0
C 1.7 1.67
D 2.4 2.28
E 3.5 3.41
(*) VALUES FOR Av AND S1 ARE FOR SOIL CLASS B
NCHRP 12-49 Seismic Hazard Proposed Practice The proposed November 2001 NCHRP 12-49 Design Earthquakes and
Performance Objectives are best described in Section 1.3 of the Recommended
LRFD Guidelines for the Seismic Design of Highway Bridges Part I
Specifications. “The USGS probabilistic maps published in 1996 (Frankel et
al., 1996) are used in formulating the design Earthquakes Response Spectrum.
The provisions provide ‘definitive performance objectives and damage states’ for
two design earthquakes with explicit design checks to ensure the performance
objectives are met. The upper-level event, termed the rare earthquake or
Maximum Considered Earthquake (MCE), describes ground motions that, for
most locations, are defined probabilistically and have a probability of
exceedance of 3% in 75 years. However, for locations close to highly active
faults, the MCE ground motions are deterministically bounded so that the
levels of ground motions do not become unreasonably high. Deterministic
bound ground motions are calculated assuming the occurrence of maximum
magnitude earthquakes on the highly active faults and are equal to 1.5 times
median ground motions for the maximum magnitude earthquake but not less
than 1.5 g for the short-period spectral acceleration plateau and 0.6g for 1.0-
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-16
second spectra acceleration. On the current MCE maps, deterministic bounds
are applied in high-seismicity portions of California, in local areas along the
California-Nevada border, along coastal Oregon and Washington, and in high-
seismicity portions of Alaska and Hawaii. In areas where deterministic bounds
are imposed, ground motions are lower than ground motions for 3% PE in 75
years. The MCE earthquake governs the limits on the inelastic deformation in
the substructures and the design displacements for the support of the
superstructure.
The lower level design event, termed the Expected Earthquake, has ground
motions corresponding to 50% PE in 75 years. This event ensures that
essentially elastic response is achieved in the substructures for the more
frequent or ‘expected earthquake’.”
According to the proposed Guidelines, “Bridges shall be designed to satisfy the
performance criteria given in Table 2-7. As a minimum, bridge shall be
designed for the life safety level of performance”.
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-17
Table 2-7: Design Earthquakes and Seismic Performance Objectives
Performance Level (1)
Probability of Exceedance
For Design Earthquake Ground Motions (4)
Life Safety
Operational
Service (2) Significant Disruption
Immediate Rare Earthquake (MCE)
3% PE in 75 years/1.5 Median Deterministic
Damage (3) Significant Minimal
Service Immediate Immediate Expected Earthquake
50% PE in 75 years Damage Minimal Minimal to None
Notes:
(1) Performance Levels:
These are defined in terms of their anticipated performance objectives in
the upper level earthquake. Life safety in the MCE event means that the
bridge should not collapse but partial or complete replacement may be
required. Since a dual level design is required the Life Safety
performance level will have immediate service and minimal damage for
the expected design earthquake. For the operational performance level
the intent is that there will be immediate service and minimal damage for
both the rare and expected earthquakes.
(2) Service Levels:
• Immediate – Full access to normal traffic shall be available following
an inspection of the bridge.
• Significant Disruption – Limited access (Reduced lanes, light
emergency traffic) may be possible after shoring, however the bridge
may need to be replaced.
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-18
(3) Damage Levels:
• None – Evidence of movement may be present but no notable
damage.
• Minimal – Some visible signs of damage. Minor inelastic response
may occur, but post-earthquake damage is limited to narrow flexural
cracking in concrete and the onset of yielding in steel. Permanent
deformations are not apparent, and any repairs could be made under
non-emergency conditions with the exception of superstructure joints.
• Significant – Although there is no collapse, permanent offsets may
occur and damage consisting of cracking, reinforcement yield, and
major spalling of concrete and extensive yielding and local buckling
of steel columns, global and local buckling of steel braces, and
cracking in the bridge deck slab at shear studs on the seismic load
path is possible. These conditions may require closure to repair the
damage. Partial or complete replacement of columns may be
required in some cases. For sites with lateral flow due to
liquefaction, significant inelastic deformation is permitted in the
piles, whereas for all other sites the foundations are capacity-
protected and no damage is anticipated. Partial or complete
replacement of the columns and piles may be necessary if significant
lateral flow occurs. If replacement of columns or other components is
to be avoided, the design approaches producing minimal or moderate
damage, such as seismic isolation or the control and reparability
design concept should be assessed.
(4) The upper-level earthquake considered in these provisions is designated
the Maximum Considered Earthquake, or MCE. In general, the ground
motions on national MCE ground motion maps have a probability of
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-19
exceedance (PE) of approximately 3% PE in 75 years. However, adjacent
to highly active faults, ground motions on MCE maps are bounded
deterministically as described above. When bounded deterministically,
MCE ground motions are lower than ground motions having 3% PE in
75 years. The performance objective for the expected earthquake is
either explicitly included as an essentially elastic design for the 50% PE
in 75 year force level or results implicitly from design for the 3% PE in
75 year force level.
The 2001 Guidelines were amended in May 2002 to delete the “Operational”
Performance Objective. The provisions were edited to reflect the consideration
of only the Life Safety Performance Objective. This change was necessary to
address the concern of some stakeholders that having more than one
performance objective as a minimum standard may create undue liability to
stakeholders that choose only a Life Safety Performance Objective with no
explicit consideration for the Operational Performance Objective. The main
changes of interest to the above-mentioned table are shown in
Table 2-8 in the Word Edit format.
Table 2-8: Design Earthquakes and Seismic Performance Objectives
Performance Level Objective(1)
Probability of Exceedance (PE)
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For Design Earthquake Ground Motions (4) Life Safety Operational
Service (2) Significant Disruption
Immediate Rare Maximum Considered Earthquake (MCE)
3% PE in 75 years/ or 1.5 Median Deterministic Damage (3) Significant Minimal
Service Immediate Immediate Expected Earthquake (EE)
50% PE in 75 years Damage Minimal Minimal to None
Notes:
(1) Performance Levels Objective:
These are defined in terms of their anticipated performance objectives in
the upper level earthquake. Life safety in the MCE event means that the
bridge should not collapse but partial or complete replacement may be
required. Since a dual level design is required the Life Safety
performance level will have immediate service and minimal damage for
the expected design earthquake. For the operational performance level
the intent is that there will be immediate service and minimal damage for
both the rare and expected earthquakes.
(2) Service Levels:
• Immediate – Full access to normal traffic shall be available following
an inspection of the bridge.
• Significant Disruption – Limited access (Reduced lanes, light
emergency traffic) may be possible after shoring, however the bridge
may need to be replaced.
(3) Damage Levels:
• None – Evidence of movement may be present but no notable
damage.
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-21
• Minimal – Some visible signs of damage. Minor inelastic response
may occur, but post-earthquake damage is limited to narrow flexural
cracking in concrete and the onset of yielding in steel. Permanent
deformations are not apparent, and any repairs could be made under
non-emergency conditions with the exception of superstructure joints.
• Significant – Although there is no collapse, permanent offsets may
occur and damage consisting of cracking, reinforcement yield, and
major spalling of concrete and extensive yielding and local buckling
of steel columns, global and local buckling of steel braces, and
cracking in the bridge deck slab at shear studs on the seismic load
path is possible. These conditions may require closure to repair the
damage. Partial or complete replacement of columns may be
required in some cases. For sites with lateral flow due to
liquefaction, significant inelastic deformation is permitted in the
piles, whereas for all other sites the foundations are capacity-
protected and no damage is anticipated. Partial or complete
replacement of the columns and piles may be necessary if significant
lateral flow occurs. If replacement of columns or other components is
to be avoided, the design approaches producing minimal or moderate
damage, such as seismic isolation or the control and repairability
design concept should be assessed.
(4) The upper-level earthquake considered in these provisions is designated
the Maximum Considered Earthquake, or MCE. In general, the ground
motions on national MCE ground motion maps have a probability of
exceedance (PE) of approximately 3% PE in 75 years. However, adjacent
to highly active faults, ground motions on MCE maps are bounded
deterministically as described above. When bounded deterministically,
MCE ground motions are lower than ground motions having 3% PE in
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-22
75 years. The performance objective for the expected earthquake is
either explicitly included as an essentially elastic design for the 50% PE
in 75 year force level or results implicitly from design for the 3% PE in
75 year force level.
SCDOT Seismic Hazard Practice The South Carolina Department of Transportation (SCDOT) has initiated the
development and implementation of a bridge seismic design program. A
central feature of the new SCDOT bridge design program is the development of
new seismic bridge design criteria and standards that: 1) incorporate a new
generation U.S. Geological Survey seismic ground shaking hazard maps, 2)
treat certain inadequacies of existing bridge design codes to adequately
address the large earthquake, and 3) address the no collapse bridge criteria
and life safety issues in the central and eastern United States. This section
summarizes the upgraded bridge seismic design provisions and describes
variations in national seismicity that motivated the development of the
SCDOT “Seismic Design Specifications for Highway Bridges”. Basically, the
revised specifications specify that the design of new bridges in South Carolina
directly account for the effects of the large earthquake as done by the State of
California. This is to ensure conformance with the guiding principle used in
the development of AASHTO provisions that the "…exposure to shaking from
the large earthquake should not cause collapse of all or part of the bridge…”
Several of the revisions were adopted from bridge design provisions of the
California Department of Transportation (Caltrans) [2], because of similar high
intensity seismic hazard at the Safety Evaluation Earthquake (SEE) level and
the state-of-practice progress gained due to recent earthquakes that have not
yet been incorporated into AASHTO.
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At least two developments of the U.S. Geological Survey during the past
several years have been a major contribution to bridge earthquake
engineering. One development was an assessment of the nature of the seismic
ground shaking hazard as it varies nationally that revealed apparent
inequalities in safety that result when a single level of probability common to
bridge code design is used. The second development was a new generation of
probabilistic ground-motion hazard maps that provide uniform hazard spectra
for exposure times of 500 and 2500 years and make possible the treatment of
the inequality in safety of bridge code design using existing earthquake
engineering design and evaluation provisions and methodology.
The new generation of probabilistic ground motion maps was produced by the
USGS under the National Earthquake Hazard Reduction Program (NEHRP)
with significant input from the committee on Seismic Hazard Maps of the
Building Seismic Safety Council (BSSC) and the Structural Engineers
Association of California (SEAOC). They allow development of uniform hazard
spectra and permit direct definition of the design spectra by mapping the
response spectral ordinates at different periods.
The recommended seismic design procedures were developed to meet current
bridge code objectives, including both serviceability and life safety in the event
of a large earthquake. The primary function of these new provisions is to
provide minimum standards for use in bridge design to maintain public safety
in the extreme earthquake likely to occur within the state of South Carolina.
They are intended to safeguard against major failures and loss of life, to
minimize damage, maintain functions, or to provide for easy repair.
For normal or essential bridges (see Table 2-9), the Single Level Design
Method is adopted by this code. This method consists of applying seismic
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-24
design loading calculated based upon the value of the spectral accelerations of
the 2%/50-year earthquake (i.e., the Safety Evaluation Earthquake).
Table 2-9: Seismic Performance Criteria
Ground Motion Level
Performance Level
Normal Bridges
Essential Bridges
Critical Bridges
Service NR* NR* Immediate Functional-Evaluation Damage NR* NR* Minimal
Service Impaired Recoverable Maintained Safety-Evaluation Damage Significant Repairable Repairable
*Functional Evaluation Not Required.
For Critical Bridges, which are designated by SCDOT, the seismic performance
goals are to be achieved by a two-level design approach, one for each of the two
earthquakes (i.e., Two-Level Design Method). In addition to the 2%/50-year
earthquake (Safety Evaluation Earthquake), critical bridges shall also be
designed to provide adequate functionality after the 10%/50-year earthquake
(Functional Evaluation Earthquake). The minimum performance levels for the
design and evaluation of bridges shall be in accordance with the level of service
and damage for the two design earthquakes as shown in Table 2-9. Service
Levels and Damage Levels are defined in these criteria. The Bridge Category
is also defined in these criteria. The SCDOT may specify project-specific or
structure-specific performance requirements different from those defined in the
table. For example, for a Critical or Essential bridge it may be desirable to
have serviceability following a 2%/50-year earthquake. The SCDOT may
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-25
require a site-specific design spectrum or a complete hazard study as part of
the design.
This new SCDOT specification establishes design and construction provisions
for bridges in South Carolina to minimize their susceptibility to damage from
earthquakes. This specification is intended to be used in conjunction with
AASHTO Division I [3] and as a replacement to Division I-A, Seismic Design,
of the same specifications. Additionally, the new specifications include
references to the AASHTO Guide Specifications for Seismic Isolation
Design [4].
The principles used for the development of the new SCDOT provisions are:
i. Small to moderate earthquakes should be resisted within the
essentially elastic range of the structural components without
significant damage. The Functional Evaluation Earthquake (FEE) is
adopted to represent seismic ground motion level produced by small
to moderate earthquakes.
ii. State-of-Practice seismic ground motion intensities and forces are
used in the design procedures.
iii. Exposure to shaking from large earthquakes should not cause
collapse of all or part of the bridge. Where possible, damage that
does occur should be readily detectible and accessible for inspection
and repair unless prohibited by the structural configuration. The
Safety Evaluation Earthquake (SEE) is adopted to represent seismic
ground motion level produced by large earthquakes.
The performance levels, expressed in terms of service levels and damage levels,
are:
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-26
(a) Service Levels
• Immediate: Full access to normal traffic is available
immediately following the earthquake.
• Maintained: Short period of closure to Public. Immediately Open
to Emergency Vehicles.
• Recoverable: Limited period of closure to Public.
• Impaired: Extended closure to Public.
(b) Damage Levels
• Minimal Damage: No collapse, essentially elastic performance.
• Repairable Damage: No collapse. Concrete cracking, spalling of
concrete cover, and minor yielding of structural steel will occur.
However, the extent of damage should be sufficiently limited that
the structure can be restored essentially to its pre-earthquake
condition without replacement of reinforcement or replacement of
structural members (i.e., ductility demands less than 4). Damage
can be repaired with a minimum risk of losing functionality.
• Significant Damage: Although there is minimum risk of collapse,
permanent offsets may occur in elements other than foundations.
Damage consisting of concrete cracking, reinforcement yielding,
major spalling of concrete, and deformations in minor bridge
components may require closure to repair. Partial or complete
demolition and replacement may be required in some cases.
Bridge structures on the state highway system are classified as “normal
bridges”, “essential bridges” or “critical bridges”. For a bridge to be classified
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-27
as an “essential bridge” or a “critical bridge”, one or more of the following items
must be present: (1) bridge is required to provide secondary life safety, (2) time
for restoration of functionality after closure creates a major economic impact,
and (3) the bridge is formally designated as critical by a local emergency plan.
Each bridge is classified as Critical, Essential or Normal as follows:
(a) Critical bridges: Bridges that must be open to all traffic once
inspected after the safety evaluation design earthquake and be
usable by emergency vehicles and for security/defense purposes
immediately after the safety evaluation design earthquake, i.e., a
2,500-year return period event.
(b) Essential bridges: Bridges that will, as a minimum, be open to
emergency vehicles and for security/defense purposes after the safety
evaluation design earthquake, i.e., a 2,500-year return period event
and open to all traffic within days after the SEE event.
(c) Normal Bridges: Any bridge not classified as a Critical or Essential
Bridge.
The SCDOT Specifications aims at including state-of-practice in seismic based
on displacement analysis for reinforced concrete components. Force reduction
factors are used for steel superstructure due to limited use of the displacement
approach in steel design of bridges. In addition, only limited level of ductility
is so far accepted for members of steel superstructure with a plate girder
system.
In September 2003, SCDOT adopted new Seismic Hazard Maps for Bridges.
These new SCDOT Seismic Hazard Maps take into account the sediment
thickness and/or the near surface weathering, updating the State seismic
1325 NCHRP 20-7(193) Task 6 Report.doc 2A-28
hazard information that was originally provided in the 2001 Specifications
based on USGS.
1325 NCHRP 20-7(193) Task 6 Report.doc 3-1
TASK 3
3 EXPAND THE EXTENT OF THE “NO ANALYSIS” ZONE
3.1 Introduction In developing the Displacement Based Approach, reference to the analysis can
be separated into two types:
a) Analysis conducted to establish seismic displacement demands on the
structures. This reference is similar to the reference made by AASHTO
Division 1-A for required seismic analysis in regions where PGA > 9% g
to determine forces. This can be referred to as “Seismic Demand
Analysis”.
b) Analysis conducted to establish the displacement capacity of the
structure, a subsystem or a component of the structure. This can be
referred to as “Seismic Capacity Analysis”. This type of analysis is also
commonly referred to as Pushover Analysis. In addition to obtaining the
displacement capacity of the structure, the “Seismic Capacity Analysis”
is used to obtain the load path and force distribution on the members of
the structure based on the hinging mechanism of these members. These
forces are used to design various members such that the developed
hinging mechanism of the overall system is confirmed.
In summary, the “Seismic Capacity Analysis” includes two parts. One is the
“Displacement Capacity” and the second is the “Capacity Design”.
1325 NCHRP 20-7(193) Task 6 Report.doc 3-2
The overall second objective identified in Task F3-4 is to increase the range of
applicability for No Analysis or Limited Analysis. This objective is made in
reference to NCHRP 12-49 Proposed Guidelines where it was found that
considerable amount of analysis was required on a larger number of bridges in
comparison with the AASHTO Division 1-A Practice. This finding is well
documented in the evaluation conducted on NCHRP 12-49 by performing trial
designed in several states. In further examining this objective, several steps
that are required to fulfill this objective are identified:
1) At a minimum, maintain the number of bridges under the “Seismic
Demand Analysis”. This objective is accomplished by comparing
Proposed Guidelines to current requirements in AASHTO Division 1-A.
2) Relative to Proposed NCHRP 12-49 Guidelines, reduce the number of
bridges where “Seismic Capacity Analysis” needs to be performed. This
objective is accomplished by identifying a threshold where implicit
procedures can be used.
3) Identify threshold where “Capacity Design” shall be used. This objective
is achieved in conjunction with the “Seismic Capacity Analysis”
requirements.
In reviewing the current State of the Practice in addressing the Range of
Applicability for No Analysis or Limited Analysis, the following sources are
examined:
1. AASHTO Division 1-A
2. Caltrans Seismic Design Criteria
3. NCHRP 12-49
4. SCDOT Specifications
1325 NCHRP 20-7(193) Task 6 Report.doc 3-3
The review of these four references is documented in Task F3-5 Report
AASHTO T-3 Support and included in Appendix 3A as background
information.
3.2 Proposed Range of Applicability of Analysis In addressing the proposed range of Applicability of Analysis, a key issue is the
selection of the most pertinent parameter indicative of the seismic demands
considered in the design of the bridge structure. The Spectral Acceleration at
1.0 second period, 1a DSS − , for the Design Spectrum is adopted considering the
following:
• 1a DSS − is a good representation of the difference in regional demands
(i.e., 1a DSS − is considerably lower in the Eastern U.S.)
• The choice of high frequency spectral indicator as recommended in
NCHRP 12-49 penalizes the Eastern U.S. for no credible justification
considering that damage to bridges is associated with low frequency
range of bridge period.
• The choice of 1a DSS − fits well with the adopted displacement approach for
bridges considering that ductility is taken into account when assessing
the capacity.
Considering the first objective of the recommended specifications addressing
the selection of a Return Period and Design Spectrum for a Single Hazard
Level pertaining to a No Collapse Criteria of bridges, the Important
Classification (IC) as defined in AASHTO Division 1-A is reduced to one
classification. Furthermore, considering the second objective of the
recommended specifications for defining and increasing the range of
1325 NCHRP 20-7(193) Task 6 Report.doc 3-4
applicability for No Analysis or Limited Analysis, the Seismic Performance
Category definition is changed to include four categories SPC A, B, C, and D
encompassing requirements for:
• Seismic Demand Analysis requirement.
• Seismic Capacity Analysis requirement.
• Capacity Design requirement.
• Level of seismic detailing requirement including four tiers corresponding
to SPC A, B, C and D.
The above-mentioned approach is an extension to the direction taken in
NCHRP 12-49 and SCDOT Specifications. The Seismic Performance
Categories SPC A, B, C and D ranges are partitioned based on the one-second
spectral acceleration 1a DSS − similarly to the SCDOT Specifications except that
the four requirements mentioned above are developed further to achieve the
second objective of the recommended specifications (i.e., reduce number of
bridges requiring analysis).
Table 3-1 shows the partition of the proposed Seismic Performance Categories
A, B, C and D.
Table 3-1: Proposed Partitions for Seismic Performance Categories A, B, C, and D
Value of 1a DSS − Importance Classification (IC)
1325 NCHRP 20-7(193) Task 6 Report.doc 3-5
1a DSS − < 0.15g A
0.15g ≤ 1a DSS − < 0.30g B 0.30g ≤ 1a DSS − < 0.50g C 0.50g ≤ 1a DSS − D
For illustration, the one-second acceleration corresponding to 0.5g, proposed to
be the threshold for SPC D, is pinpointed in Figures 3-1 and 3-2, showing the
acceleration response spectrum for Magnitude 6.5 and 8 (Type D Soil),
respectively. As seen from the figures, SPC D would include sites with Peak
Bedrock Acceleration greater than 0.3g for magnitude 6.5 and greater than
0.2g for magnitude 8. This shows that SPC D is rather conservative in
applying the most stringent criteria on the sites mentioned above.
Tables 3-2 thru 3-4 show the one-second spectral acceleration for Caltrans SDC
Magnitude 6.5, 7 and 8 for Soil Type B thru E. The numbers that are not
shaded represent values greater than the 0.5g threshold considered for SPC D.
Table 3-5 shows the one-second acceleration for (Division 1A) design spectrum
for Soil Type 1 thru 4. The numbers that are not shaded represent values
greater than 0.5g considered the threshold for SPC D.
Table 3-6 shows the one-second acceleration modified for Type B and D soil for
the sites selected in Task 2. Each site is assigned an SPC based on the
proposed partition shown in Table 3-1. Table 3-6 reflects the distribution of
SPC A, B, C and D given Type B and D soils.
1325 NCHRP 20-7(193) Task 6 Report.doc 3-6
Figure 3-1: Elastic Response Spectra Curves (5% Damping) for Soil Profile Type D (M = 6.5 ± 0.25) (Caltrans SDC)
1325 NCHRP 20-7(193) Task 6 Report.doc 3-7
Figure 3-2: Elastic Response Spectra Curve (5% Damping) for Soil Profile Type D (M = 8.0 ± 0.25) (Caltrans SDC)
1325 NCHRP 20-7(193) Task 6 Report.doc 3-8
Table 3-2: 1.0 sec. Spectral Acceleration for Magnitude 6 "A"(g) soil B soil C soil D soil E
0.1 0.08 0.14 0.19 0.280.2 0.16 0.21 0.32 0.520.3 0.24 0.37 0.45 0.690.4 0.35 0.48 0.55 0.800.5 0.41 0.53 0.610.6 0.48 0.61 0.70
Table 3-3: 1.0 sec. Spectral Acceleration for Magnitude 7 "A"(g) soil B soil C soil D soil E
0.1 0.10 0.17 0.24 0.360.2 0.21 0.32 0.41 0.690.3 0.29 0.44 0.54 0.810.4 0.42 0.59 0.66 0.990.5 0.52 0.68 0.770.6 0.67 0.86 1.000.7 0.92 1.19 1.38
Table 3-4: 1.0 sec. Spectral Acceleration for Magnitude 8 "A"(g) soil B soil C soil D soil E
0.1 0.12 0.21 0.29 0.430.2 0.24 0.38 0.48 0.750.3 0.35 0.53 0.64 0.930.4 0.45 0.63 0.71 1.020.5 0.57 0.73 0.900.6 0.72 0.93 1.120.7 0.96 1.25 1.45
Table 3-5: 1.0 sec. Spectral Acceleration (Division 1A)
"A"(g) soil 1 soil 2 soil 3 soil 40.1 0.12 0.14 0.18 0.240.2 0.24 0.29 0.36 0.480.3 0.36 0.43 0.54 0.720.4 0.48 0.58 0.72 0.960.5 0.60 0.72 0.90 1.20.6 0.72 0.86 1.08 1.440.7 0.84 1.01 1.26 1.68
1325 NCHRP 20-7(193) Task 6 Report.doc 3-9
Table 3-6: Seismic Performance Category for Selected Sites
State CITY
Type B Soil5%/50 yr
Sa1, g SPC
Type D Soil5%/50 yr
Sa1, g SPCCA Daly City 1.12 D 1.67 DCA San Francisco 0.69 D 1.04 DCA SFOBB 0.62 D 0.93 DCA Berkeley 0.83 D 1.24 DCA Benicia Martinez 0.55 D 0.83 DCA Los Angeles 0.54 D 0.81 DCA Vincent Thomas 0.56 D 0.84 DCA Long Beach 0.49 C 0.73 DCA Coronado Bridge 0.34 C 0.58 DWA Seattle 0.33 C 0.58 DWA Tacoma North 0.30 C 0.54 DUT Salt Lake City 0.42 C 0.67 DCA Salt Lake City 0.42 C 0.66 DIN Evansville 0.11 A 0.26 B
MO St. Louis 0.10 A 0.24 BKY Paducah 0.24 B 0.46 CTN Union City 0.27 B 0.50 DTN Memphis 0.19 B 0.39 CTN Memphis 0.20 B 0.41 CSC Charleston 0.15 B 0.34 CAZ Phoenix 0.04 A 0.09 A
The three requirements for each of the proposed Seismic Performance
Categories are as follows:
1. SPC A
a. No Displacement Capacity Check Needed
b. No Capacity Design Required
c. Tier I No Detailing Requirements
1325 NCHRP 20-7(193) Task 6 Report.doc 3-10
2. SPC B
a. Implicit Displacement Capacity Check Required (i.e., use a Closed
Form Solution Formula)
b. No Capacity Design Required
c. Tier II Level of Detailing
3. SPC C
a. Implicit Displacement Capacity Check Required
b. Capacity Design Required
c. Tier III Level of Detailing
4. SPC D
a. Pushover Analysis Required
b. Capacity Design Required
c. Tier IV Level of Detailing
The level of detailing for Tiers I, II, III, and IV will consider at a minimum the
following:
• Column Longitudinal Reinforcement Splicing
• Column Transverse Reinforcement Splicing
• Column Plastic Hinge Zone Identification
• Joint Shear Reinforcement
1325 NCHRP 20-7(193) Task 6 Report.doc 3-11
• Knee Joint Reinforcement
• Bent Cap Continuous Reinforcement
• Superstructure Continuous Reinforcement
• Footing Shear Reinforcement
• Bent Cap Shear Reinforcement
• Plate Girder Bracing/Diaphragm Detailing
The three requirements for each of SPC A, B, C and D will follow the core
flowchart that was presented in Task F3-5 and shown in Figure 3-3.
Figure 3-3: Core Flowchart
SPC "A"
D em an dA n aly sisSPC "B " Im plic it
C apacity
Y es
No
Y es1D
C ≤ T ier IID etailin g
C om pleteYes
SPC "C"
No No
Y es1D
C ≤ C apacityD esig n
T ier IIID etailin g
Yes
SPC "D"
No No
P u sh overC apacityA n alys is
1DC ≤ T ier IV
D etailin g
Yes
No
A d ju st B rid g eC h aracteristicsDep en d s on Ad ju stm en ts
Yes
D em an dA n aly sis
Im plic itC apacity C om plete
C om plete
D em an dA n aly sis
C apacityD esig n C om plete
1325 NCHRP 20-7(193) Task 6 Report.doc 3-12
The major performance measures that govern the design for each of the
Seismic Performance Categories include the following:
1. Column Shear requirement
2. Drift Capacity requirement
3. Seat Width requirement
3.3.1 Column Shear Requirement for SPC B The shear demand for a column, Vd, in SPC B shall be determined based on the
lesser of:
• The force obtained from an elastic linear analysis
• The force corresponding to plastic hinging of the column
The column shear strength capacity shall be calculated based on
n oV Vφ ≥ 0.85φ = (3.1)
n c sV V V= + (3.2)
,v yhs
A f DV
s⎛ ⎞
= ⎜ ⎟⎝ ⎠
where 2v bA n Aπ⎛ ⎞= ∗ ∗⎜ ⎟⎝ ⎠ (3.3)
n = number of individual interlocking spiral or hoop core sections.
For tied columns or pier walls (in the weak direction).
1325 NCHRP 20-7(193) Task 6 Report.doc 3-13
v yhs
A f DV
s⎛ ⎞
= ⎜ ⎟⎝ ⎠
(3.4)
vA = Total area of the shear reinforcement
c c eV v A= × (3.5)
0.8e gA A= × (3.6)
1 3.52000c c c
g
Pv f fA
α⎛ ⎞
′ ′ ′= + ≤⎜ ⎟⎜ ⎟⎝ ⎠
(3.7)
Spirally reinforced columns 0.015 s ytfα ρ′ = (3.8)
Rectangular hoop reinforced columns 0.030 w ytfα ρ′ = (3.9)
Where the spiral reinforcement ratio,
4 sps
ADs
ρ = (3.10)
and the web reinforcement ratio
vw
Abs
ρ = (3.11)
1325 NCHRP 20-7(193) Task 6 Report.doc 3-14
3.3.2 Column Shear Requirement for SPC C The shear demand for a column, Vd, in SPC B shall be determined based on the
force corresponding to plastic hinging of the column including an overstrength
factor
The column shear strength capacity shall be calculated based on
n oV Vφ ≥ 0.85φ = (3.12)
n c sV V V= + (3.13)
,v yhs
A f DV
s⎛ ⎞
= ⎜ ⎟⎝ ⎠
where 2v bA n Aπ⎛ ⎞= ∗ ∗⎜ ⎟⎝ ⎠
(3.14)
n = number of individual interlocking spiral or hoop core sections.
For tied columns or pier walls (in the weak direction).
v yhs
A f DV
s⎛ ⎞
= ⎜ ⎟⎝ ⎠
(3.15)
vA = Total area of the shear reinforcement
c c eV v A= × (3.16)
0.8e gA A= × (3.17)
1325 NCHRP 20-7(193) Task 6 Report.doc 3-15
1 3.52000c c c
g
Pv f fA
α⎛ ⎞
′ ′ ′= + ≤⎜ ⎟⎜ ⎟⎝ ⎠
(3.18)
Spirally reinforced columns 0.010 s ytfα ρ′ = (3.19)
Rectangular hoop reinforced columns 0.020 w ytfα ρ′ = (3.20)
Where the spiral reinforcement ratio,
4 sps
ADs
ρ = (3.21)
and the web reinforcement ratio
vw
Abs
ρ = (3.22)
3.4 Drift Capacity for SPC B and SPC C Columns for bridges in SPC B are targeted for a limited drift corresponding to
minor damage. Columns for bridges in SPC C are targeted for a maximum
drift corresponding to moderate damage. The approach taken to come up with
a closed form solution is to equally weigh in the results of numerical methods
as well as experimental testing of various columns.
Considering the numerical approach as described below, columns with
diameter ranging from 3 feet to 7 feet having 1 to 4% longitudinal
reinforcement and height ranging between 20 to 50 feet are considered. The
1325 NCHRP 20-7(193) Task 6 Report.doc 3-16
different permutations are shown in Table 3-7. A regression analysis is
performed and a lower bound curve is identified in Figure 3-4 for the following:
a. Curve 1, designated as C1, showing drift capacity corresponding to
column yielding
b. Curve 2, designated as C2, showing drift capacity corresponding to
concrete spalling
c. Curve 3, designated as C3, showing drift capacity corresponding to a
column ductility of 4.
The drift capacity for all three curves are shown as a function of the
slenderness ratio FbL where:
F = Flixity Factor ranging from 1 to 2
b = column width or diameter
L = column clear height
Experimental results are considered based on the statistical study adopted by
Pacific Earthquake Engineering Research Center (PEER) and reported by
Berry and Eberhand August 2003. The recommended equation by Berry and
Eberhard at the onset of spalling is:
11.6 1 110( )g c
PDH A f H
⎛ ⎞⎛ ⎞Δ ⎜ ⎟= − +⎜ ⎟⎜ ⎟′ ⎜ ⎟⎝ ⎠⎝ ⎠ (3.23)
Considering a typical bridge column axial load corresponding to 0.1 g cA f ′ , Curve
4, designated as C4, shows graphically the recommended equation by Berry
and Eberhand (PEER).
1325 NCHRP 20-7(193) Task 6 Report.doc 3-17
The recommended maximum drift capacity for SPC B is further defined as
Curve 5, designated as C5, and shown in Figure 3-4.
Curve 2 + Curve 4Curve 52
= (3.24)
The recommended maximum drift capacity for SPC C is further defined as
Curve 6, designated as C6, and shown in Figure 3-4.
Curve 3 + Curve 4Curve 62
= (3.25)
Table 3-7: Column Parameters
Column Diameter D (ft) ρ
(%) Column Height L (ft)
3 1,2,3,4 20 4 1,2,3,4 20,30 5 1,2,3,4 20,30,40 6 1,2,3,4 30,40,50 7 1,2,3,4 30,40,50
1325 NCHRP 20-7(193) Task 6 Report.doc 3-18
0.00
1.00
2.00
3.00
4.00
5.00
6.00
0.1 0.15 0.2 0.25 0.3
Fb/L
Drif
t Cap
acity
(%) Yield (C1)
Spalling (C2)Ductility 4 (C3)Experimental (C4)SPC B (C5)SPC C (C6)
Figure 3-4: Proposed Drift Capacity for SPC B and C
3.5 Hinge Seat Requirement The calculation for a hinge seat width involves four components:
a. Minimum edge distance
b. Other movement attributed to prestress shortening, creep, shrinkage,
and thermal expansion or contraction
c. Skew effect
d. Relative hinge displacement
3.5.1 Minimum Edge Distance The minimum edge distance set by Division IA and NCHRP 12-49 is set at 4
inches. It is recommended to retain this value.
SPC C
SPC B
1325 NCHRP 20-7(193) Task 6 Report.doc 3-19
3.5.2 Other Movement Division IA currently has a movement rating of 2 inches per 100 feet for SPC B
and a movement rating of 3 inches per 100 feet for SPC C & D.
The seat width based on NCHRP 12-49 is calculated as:
21(1 1.25 )0.10 0.0017 0.007 0.05 1 2 vF SBN L H H
L cos α
⎡ ⎤ +⎛ ⎞⎢ ⎥= + + + × + ⎜ ⎟⎢ ⎥⎝ ⎠⎣ ⎦ (3.26)
L = the distance between joints in meters
H = the tallest pier between the joints in meters
B = the width of the superstructure in meters
α = the skew angle
The term .0017L equates to 100.0017 1003.3m
ft⎛ ⎞⎜ ⎟⎝ ⎠
or .05 100m
ft equal to 2 inches
per 100 feet.
Three alternatives are considered for including “other” movement in the seat
width equation:
a. The first alternative considers the temperature movement tΔ and other
movements as calculated for various states based on their extreme
temperature range, in addition to prestress and shortening, and thermal
expansion or contraction.
b. The second alternative is a 2-inch movement per 100 feet, which is quite
conservative.
1325 NCHRP 20-7(193) Task 6 Report.doc 3-20
c. The last alternative has a 1-inch movement rating per 100 feet
considered an average nominal value in Practice especially in
combination with seismic movement.
It is recommended for clarity and transparency to adopt Alternative (a) stated
above.
3.5.3 Skew Effect A comparison of Equation 6.3.1 adopted in NCHRP 12-49 to Division I-A seat
width magnification for various skew angles is shown in Figure 3-5. As seen
from Figure 3-5, NCHRP 12-49 magnification is larger than Division I-A.
Doubling the magnification set in Division 1A as shown in the upper bound
curve as ( )21 4000
S+ is recommended. This recommendation is based on the
failures observed in past earthquakes for bridges with skewed bents.
1
1.05
1.1
1.15
1.2
1.25
1.3
1.35
1.4
1.45
0 5 10 15 20 25 30 35 40 45
Skew Angle
Am
plifi
catio
n Fa
ctor NCHRP 12-49
Division 1AProposed
Figure 3-5: Skew Effect Seat Width Amplification
Factor for Various Skew Angles
1325 NCHRP 20-7(193) Task 6 Report.doc 3-21
3.5.4 Relative Hinge Displacement The relative hinge displacement, Deq, is determined following guidelines by
Desroches & Fenves adopted by the recently published FHWA “Seismic
Retrofitting Manual for Highway Structures, 2004.”
( )2 2min max 12 min max2eqD D D D Dρ= + − (3.27)
where, Dmin = Displacement of the short period frame
Dmax = Displacement of the long period frame.
The correlation coefficient 12ρ is calculated as:
( )( )( ) ( ) ( )
3/ 22
12 2 22 2
8 1
1 4 1
ε β βρ
β ε β β
+=
− + + (3.28)
where 2
1
TT
β = T2 and T1 being the first and second modes of the structure
system.
The damping ε is calculated as:
( )15% 1 0.95 0.05ε μ μπ
= + − − (3.29)
where μ is the ductility factor
Consider the displacement ratio α :
min
max
DD
α = (3.30)
1325 NCHRP 20-7(193) Task 6 Report.doc 3-22
( )2 2 2 2max max 12 max2eqD D D Dα ρ α= + − (3.31)
( )2121 2α αρ= + − (3.32)
In the long period range, α , is also equal to the ratio of the short period frame
over the long period frame.
short
long
TT
α = (3.33)
Figure 3-6 shows Dmax vs. the ratio α for the following:
a. Deq for a target ductility of 2 shown as Curve 1
b. Deq for a target ductility of 4 shown as Curve 2
c. Caltrans SDC shown as Curve 3
d. Relative hinge displacement based on (Trocholak is et. al. 1997) shown
as Curve 4
Considering that a variation from the design plans of the structure cannot be
avoided during the life of the structure and that a substantial drop in the
required seat width is only achieved for an α greater than 0.8, it is
recommended that:
• Deq is equal to 1.1 Dmax the peak value of Curves 1 and 2.
Furthermore, a safety factor of 1.5 is proposed for regions other than California
as described in Task 2.
• Deq is equal to 1.1 Dmax for California
1325 NCHRP 20-7(193) Task 6 Report.doc 3-23
• Deq is equal to 1.65 Dmax for states other than California
0.00
0.20
0.40
0.60
0.80
1.00
1.20
1.40
1.60
0 0.2 0.4 0.6 0.8 1 1.2
Ratio of Tshort/Tlong
Rat
io o
f Deq
/Dm
ax
Curve 1Curve 2Curve 3Curve 4
Figure 3-6: Relative Seismic Displacement vs. Period Ratio
The proposed seat width requirement is compared to NCHRP 12-49 Equation
3.26 (shown in Curves 1 and 2 of Figure 3-7) and Division 1A seat width
requirement (shown in Curves 3 and 4 of Figure 3-7). The following is
considered for Equation 3.26:
a) substitution of SDR 2 “FvS1” by the maximum value of 0.25.
b) substitution of “B/L” ratio by the maximum value of 3/8.
c) Substitution of SDR 3 “FvS1” by the maximum value of 0.40.
The proposed seat width requirement is shown in Figures 3-7 and 3-8 for
H = 20 ft and 30 ft, respectively independent of any skew effect. The proposed
seat width requirement is illustrated with four cases identified in four curves:
1325 NCHRP 20-7(193) Task 6 Report.doc 3-24
1. Curve 5 for FvS1 = 0.15g corresponding to a period of 1 second for the
flexible frame with Deq equal to 1.65 Dmax.
2. Curve 6 for FvS1 = 0.5g corresponding to a period of 1 second for the
flexible frame with Deq equal to 1.65 Dmax.
3. Curve 7 for FvS1 = 0.15g corresponding to a period of 2 second for the
flexible frame with Deq equal to 1.65 Dmax.
4. Curve 8 for FvS1 = 0.5g corresponding to a period of 2 second for the
flexible frame with Deq equal to 1.65 Dmax.
The calculation for seat width requirement of the four cases above considers a
1-inch displacement per 100 feet for displacement other than seismic. The
choice of one inch per 100 feet leads to shallower slope of lines 5 thru 8 and
reinforces the choice of a realistic TΔ rather than the conservative 2 inches per
100 feet adopted in NCHRP 12-49. It is expected that the choice of alternative
(a) identified in Section 3.5.2 would yield a movement not exceeding one inch
per 100 feet of bridge length. The choice of a realistic eqΔ is important for the
design of hinges within-a span and the selection of a reasonable dimension for
the bent cap width.
1325 NCHRP 20-7(193) Task 6 Report.doc 3-25
0
5
10
15
20
25
30
35
40
45
0 200 400 600 800 1000
Bridge Length (ft.)
Seat
Wid
th (i
n.)
Curve 1 SDR 2Curve 2 SDR 3Curve 3 SPC BCurve 4 SPC C&DCurve 5 .15g, 1 secCurve 6 .5g, 1 secCurve 7 .15g, 2 secCurve 8 .50g, 2 sec
Figure 3-7: Proposed Seat Width Compared to NCHRP 12-49
and DIV 1A (H=20ft)
05
1015
2025
3035
40
45
50
0 200 400 600 800 1000Bridge Length (ft.)
Seat
Wid
th (i
n.)
Curve 1 SDR 2Curve 2 SDR 3Curve 3 SPC BCurve 4 SPC C&DCurve 5 .15g, 1 secCurve 6 .5g, 1 secCurve 7 .15g, 2 secCurve 8 .50g, 2 sec
Figure 3-8: Proposed Seat Width Compared to NCHRP 12-49
and DIV 1A (H=30ft)
1325 NCHRP 20-7(193) Task 6 Report.doc 3-26
REFERENCE
Berry, Michael and Eberhand, Marc, Pacific Earthquake Engineering Research
Center (PEER), “Estimating Flexural Damage in Reinforced Concrete
Columns,” University of California, Berkeley, August 2003.
DesRoches, Reginald, and Gregory Fenves, New design and analysis
procedures for intermediate hinges in multiple-frame bridges. Berkeley, Calif.:
Earthquake Engineering Research Center, University of California. 1997.
202p. (UCB/EERC 97/12).
Aschheim, Mark, and Jack P. Moehle, Shear Strength and Deformability of RC
Bridge Columns Subjected to Inelastic Cyclic Displacements, Berkeley, Calif.:
Earthquake Engineering Research Center, University of California. March
1992. (UCB/EERC 92/04).
1325 NCHRP 20-7(193) Task 6 Report.doc 3A-1
TASK 3
APPENDIX 3A
1325 NCHRP 20-7(193) Task 6 Report.doc 3A-2
AASHTO Division 1-A Range of Applicability of Analysis The analysis requirements based on AASHTO Division 1-A are derived based
on the Seismic Performance Category (SPC) and the regularity or irregularity
of a given bridge. These requirements are relevant to the “Seismic Demand
Analysis” mentioned above.
Each bridge is assigned to one of four Seismic Performance Categories (SPC), A
through D, based on the Acceleration Coefficient (A) and the Importance
Classification (IC), as shown in Table 3-1. Minimum analysis and design
requirements are governed by the SPC.
Table 3-1: Seismic Performance Category (SPC)
Acceleration Coefficient
Importance Classification (IC)
A I II A ≤ 0.09 A A 0.09 < A ≤ 0.19 B B 019 < A ≤ 0.29 C C 0.29 < A D C
An Importance Classification (IC) is assigned for all bridges with an
Acceleration Coefficient greater than 0.09 for the purpose of determining the
Seismic Performance Category (SPC) as follows:
1. Essential bridges – IC = I
2. Other Bridges – IC = II
Bridges shall be classified on the basis of Social/Survival and Security/Defense
requirements, guidelines for which are given in the Commentary of AASHTO
Division 1-A.
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Minimum requirements for the selection of an analysis method for a particular
bridge type are given in Table 3-2. Applicability is determined by the
“regularity” of a bridge which is a function of the number of spans and the
distribution of weight and stiffness. Regular bridges have less than seven
spans, no abrupt or unusual changes in weight, stiffness, or geometry and no
large changes in these parameters from span-to-span or support-to-support
(abutments excluded). They are defined in Table 3-3. Any bridge not
satisfying the requirements of Table 3-3 is considered to be “not regular”. A
more rigorous, generally accepted analysis procedure may be used in lieu of the
recommended minimum such as the Time History Method (Procedure 4).
Table 3-2: Minimum Analysis Requirements
Seismic Performance
Category
Regular Bridges with
2 Through 6 Spans
Not Regular Bridges with
2 or More Spans A Not Required Not Required
Use Procedure Use Procedure B, C, D 1 or 2 3
Procedure 1. Uniform Load Method
Procedure 2. Single-Mode Spectral Method
Procedure 3. Multimode Spectral Method
Procedure 4. Time History Method
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Table 3-3: Regular Bridge Requirements Parameters
Value Number of Spans 2
3 4 5 6
Maximum subtended angle (curved bridge)
90º 90º 90º 90º 90º
Maximum span length ratio from span-to-span
3 2 1.5 1.5
Maximum bent/pier stiffness ratio from span-to-span (excluding abutments)
- 4 4 3 2
Note: All ratios expressed in terms of the smaller value.
Curved bridges comprised of multiple simple spans shall be considered to be
“not regular” bridges if the subtended angle in plan is greater than 20º; such
bridges shall be analyzed by either Procedure 3 or 4.
Caltrans Range of Applicability of Analysis The Caltrans Seismic Design Criteria (SDC) specify the minimum seismic
design requirements that are necessary to meet the performance goals
established for Ordinary Standard bridges.
A structure must meet all of the following requirements to be classified as an
Ordinary Standard bridge:
• Span lengths less than 300 feet (90 m).
• Constructed with normal weight concrete girder, and column or pier
elements.
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• Horizontal members either rigidly connected, pin connected, or
supported on conventional bearings by the substructure, isolation
bearings and dampers are considered nonstandard components.
• Dropped bent caps or integral bent caps terminating inside the exterior
girder, C-bents, outrigger bents, and offset columns are nonstandard
components.
• Foundations supported on spread footing, pile cap with piles, or pile
shafts.
• Soil that is not susceptible to liquefaction, lateral spreading, or scour.
Ordinary Nonstandard bridges require project specific criteria to address their
non-standard features.
Based on Caltrans SDC, each bridge presents a unique set of design
challenges. The designer is given the latitude to determine the appropriate
methods and level of refinement necessary to design and analyze each bridge
on a case-by-case basis. Situations may arise that warrant detailed attention
beyond what is provided in the SDC. The designer is referred to other
resources to establish the correct course of action. The Senior Seismic
Specialists, the Earthquake Committee, and the Earthquake Engineering
Branch of the Office of Earthquake Engineering and Design Support should be
consulted for recommendations.
The global displacement demand estimate for Ordinary Standard bridges is
determined by linear elastic response spectrum analysis utilizing effective
section properties.
Equivalent Static Analysis is used to determine the displacement demand if a
dynamic analysis will not add significantly more insight into behavior. The
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Equivalent Static Analysis is best suited for bridges or individual frames with
the following characteristics:
• Response primarily captured by the fundamental mode of vibration with
uniform translation
• Simply defined lateral force distribution (e.g., balanced spans,
approximately equal bent stiffness)
• Low skew
Elastic Dynamic Analysis is used to determine the displacement demand for all
other Ordinary Standard bridges.
The global displacement demand estimate shall include the effects of
soil/foundation flexibility if they are significant.
Following the Caltrans Seismic Design Criteria V1.2 the Inelastic Static
Analysis commonly referred to as “push over” analysis is to be used to
determine the reliable displacement capacities of a structure or frame as it
reaches its limit of structural stability.
The two-dimensional plane frame “push over” analysis of a bent or frame can
be simplified to a column model (fixed-fixed or fixed-pinned) if it does not cause
a significant loss in accuracy in estimating the displacement demands or the
displacement capacities. The effect of overturning on the column axial load
and associated member capacities must be considered in the simplified model.
Simplifying the demand and capacity models is not permitted if the structure
does not meet the following stiffness and period requirements:
a) Balanced Stiffness
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Constant Width Frame Variable Width Frame
Stiffness ratio between any two bents within a frame or between any two columns within a bent.
0.5ei
ej
kk ≥
0.5
ei
iej
j
km
km
≥
Stiffness ratio between adjacent bents within a frame or between adjacent columns within a bent.
0.75ei
ej
kk ≥
0.75
ei
iej
j
km
km
≥
eik = The smaller effective bent or column stiffness im = Tributary mass of column or bent i
ejk = The larger effective bent or column stiffness jm = Tributary mass of column bent j
b) Balanced Frame Geometry
The ratio of fundamental periods of vibration for adjacent frames in the
longitudinal and transverse direction shall satisfy:
0.7i
j
TT ≥ where
iT = Natural period of the less flexible frame
jT = Natural period of the more flexible frame
In addition to the global analysis conducted on the overall structure to
determine displacement demands, a Stand-Alone analysis (i.e., shake down) is
performed in both the transverse and longitudinal directions. This analysis is
performed on individual frames that are separated by a superstructure
expansion joint.
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In summary, Caltrans SDC v1.2 gives some latitude to the bridge engineer to
decide the type and amount of analysis to be conducted. This latitude is offset
by a quality control mechanism that is established in Caltrans and may not
exist in all other agencies nationwide. Furthermore, as Caltrans uses one set
of standard details for the whole bridge inventory, bridges in lower seismic
zones may end up with more stringent requirements and detailing that are not
needed in lower seismic zones. Therefore, in examining the Caltrans Practice,
it is deemed important to recognize that this Practice needs to be selectively
replicated for use by other states or agencies.
NCHRP 12-49 Range of Applicability of Analysis Each bridge is assigned a Seismic Hazard Level that is the highest level
determined by the valued of FvS1 or FaSs from Table 3-4 for the MCE event.
The spectral acceleration ordinates FvS1 and FaSs are illustrated in Figure 3-1.
Table 3-4: Seismic Hazard Levels Seismic Hazard Level
Value of FvS1 Value of FaSs
I FvS1≤0.15 FaSs≤0.15 II 0.15 < FvS1≤0.25 0.15 < FaSs≤0.35 III 0.25 < FvS1≤0.40 0.35 < FaSs≤0.60 IV 0.40 < FvS1 0.60 < FaSs
Notes:
1. For the purposes of determining the Seismic Hazard Level for Site Class
E Soils, the value of Fv and Fa need not be taken larger than 2.4 and 1.6
respectively when S1 is less than or equal to 0.10 and Ss is less than
0.25.
2. For the purposes of determining the Seismic Hazard Level for Site Class
F Soils, Fv and Fa values for Site Class E soils may be used with the
adjustment described in Note 1 above.
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Figure 3-1: Design Spectrum
Each bridge is designed, analyzed and detailed for seismic effects in accordance
with Table 3-5. Seismic Design and Analysis Procedures (SDAP) are described
in Section 4 of the NCHRP 12-49 document. Minimum seismic design
requirements (SDR) for SDR 1 and 2, SDR 3 and SDR 4 are given in Sections 6,
7 and 8, respectively of NCHRP 12-49.
Table 3-5: Seismic Design and Analysis Procedures (SDAP) and Seismic Design Requirements (SDR)
Life Safety Seismic Hazard Level SDAP SDR
I A1 1 II A2 2 III B/C/D/E 3 IV C/D/E 4
SDAP A1 and A2 do not have dynamic analysis requirements. Bridges
qualifying for SDAP B do not require a seismic demand analysis but capacity
design principles and minimum design details are required. SDAP C is the
Capacity Spectrum Design Method. SDAP C combines a demand and capacity
Site Class Spectrum
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analysis. The procedure applies only to bridges that behave essentially as a
single degree-of-freedom system. SDAP C is restricted to bridges with a very
regular configuration provided the abutments are not considered part of the
Earthquake Resistant System.
SDAP D is the Elastic Response Spectrum Method. SDAP D is a one-step
design procedure using an elastic (cracked section properties) analysis. Either
the Uniform Load or Multimode method of analysis may be used. The analysis
shall be performed for the 3% PE in 75-year/1.5 mean deterministic and the R-
Factors given in Tables 3-6 and 3-7. Capacity design principles shall be used
for column shear design and the design of all column connections and
foundation design. If sacrificial elements are part of the design (i.e., shear
keys) they shall be sized to resist the 50% PE in 75-year forces and the bridge
shall be capable of resisting the 3% PE in 75-year/1.5 mean deterministic
forces without the sacrificial elements (i.e., two analyses are required if
sacrificial elements exist in a bridge).
SDAP E is the Elastic Response Spectrum Method with Displacement Capacity
Verification. SDAP E requires an elastic (cracked section properties) response
spectrum analysis for the governing design spectra (50% PE in 75-year or 3%
PE in 75-year/1.5 mean deterministic) and P-Δ design check. The results of
these analyses shall be used to perform preliminary flexural design of plastic
hinges in columns and to determine the displacement of the structure. To take
advantage of the higher R-Factors in Table 3-6, displacement capacities shall
be verified using two-dimensional nonlinear static (pushover) analyses in the
principal structural directions. Design forces on substructure elements may be
reduced below those obtained for the 3% PE in 75-year event/1.5 mean
deterministic divided by the R-Factor. If sufficient displacement capacity
exists, the substructure design forces may be further reduced an additional
30% for a new sizing of the substructure members provided a second
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displacement capacity is performed. Capacity design principles shall be used
to design the connection of the columns to the superstructure and foundation
and for column shear design.
Table 3-6: Base Response Modification Factors, RB, for Substructure Performance
Objective Life Safety Substructure Element
SDAP D SDAP E Wall Piers – larger dimension 2 3 Columns – Single and Multiple 4 6 Pile Bents and Drilled Shafts – Vertical Piles – above ground 4 6
Pile Bents and Drilled Shafts – Vertical Piles – 2 diameters below ground level – No owners approval required
1 1.5
Pile Bents and Drilled Shafts – Vertical Piles – in ground – Owners approval required. N/A 2.5
Pile Bents with Batter Piles N/A 2 Seismically Isolated Structures 1.5 1.5 Steel Braced Frame – Ductile Components 3 4.5 Steel Braced Frame – Nominally Ductile Components 1.5 2 All Elements for Expected Earthquake 1.3 1.3
Table 3-7: Response Modification Factors, R – Connections
Connection All Performance Objectives
Superstructure to abutment 0.8 Expansion joints within a span of the superstructure 0.8 Columns, piers, or pile bents to cap beam or superstructure 0.8
Columns or piers to foundations 0.8
Following the NCHRP 12-49 specifications, the displacement capacity
verification analysis shall be applied to individual piers or bents to determine
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the lateral load-displacement behavior of the pier or bent. The capacity
evaluation shall be performed for individual piers or bents in the longitudinal
and transverse direction separately.
The capacity evaluation shall identify the component in the pier or bent that
first reaches its inelastic deformation capacity. The displacement at which the
first component reaches its maximum permitted deformation capacity defines
the maximum displacement capacity, Δcapacity for the pier or bent and this shall
exceed the factored displacement demand, Δ, according to the following
requirement:
1.5Δ ≤ Δ capacity
The model for the displacement capacity verification is based on nominal
capacities of the inelastic components. Stiffness and strength degradation of
inelastic components and effects of loads acting through the lateral
displacement shall be considered.
In examining SDAP E, which is based on a force reduction approach with
higher Response Modification factors RB provided a displacement verification is
performed, it is deemed important to reiterate the following:
a) NCHRP 12-49 recognizes a 30% further reduction of substructure design
forces provided a displacement capacity is performed. This statement is
in tune with current state of the practice highlighting the advantages of
using a displacement approach.
b) The displacement capacity is established based on the weakest
component; therefore no strength loss or degradation is considered
acceptable. Even though this practice can be adopted for simplicity, it is
considered extremely conservative when it is associated with the
following:
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• The use of 1.5 factor for displacement demands established based
on the 3% PE in 75-year/1.5 mean deterministic event.
• The use of nominal properties for establishing capacities of
inelastic components.
As seen above, the NCHRP 12-49 recognizes the advantage of using a
displacement approach (a) but then retracts or offsets this advantage by
placing a 1.5 factor on the displacement demand.
In retrospect, the adoption of displacement capacity determination based on
the weakest component (i.e. no strength loss) is consistent with the state of the
practice aiming for some degree of simplicity in performing the push over
analysis. Furthermore, the use of nominal properties is also consistent with
current state of the practice.
In summary, the use of a 1.5 factor for displacement demand is considered
excessive and unwarranted considering the inherent conservatism in
establishing the displacement capacity.
SCDOT Specifications Range of Applicability of Analysis Similar to AASHTO, Division 1-A, the “Seismic Demand Analysis”
requirements in the SCDOT Specifications are derived based on the Seismic
Performance Category (SPC) and the regularity or irregularity of a given
bridge. The regularity requirements in the SCDOT Specifications are
identical to those from AASHTO Division 1-A Specifications.
The seismic hazard varies form very small to high across the State of South
Carolina. Therefore, for purposes of design, four Seismic Performance
Categories (SPC) are defined on the basis of the spectral acceleration for the
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one second period of the 2%/50-year earthquake, SD1-SEE, and the Importance
Classification (IC) as shown in Table 3-6. The design response spectral
acceleration at 1.0-second period D1-SEES is shown in Figure 3-2. Different
degrees of complexity and sophistication of seismic analysis and design are
specified for each of the four Seismic Performance Categories.
Table 3-6: Seismic Performance Category (SPC)
Importance Classification (IC) Value of Spectral Acceleration, SD1-
SEE I II III
SD1-SEE<0.30g B B A 0.3g≤SD1-SEE<0.45g C C B 0.45g≤ SD1-SEE<0.6g D C C 0.6g≤ SD1-SEE D D C
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Figure 3-2: Design response spectrum curve
The design spectrum for the FEE (10% in 50 years) and the SEE (2% in 50
years) were developed using the 1997 NEHRP Maps. The curves are anchored
to the 0.2 second mapped design spectral accelerations for Site Class B rock site. As shown in Figures 3-3 and 3-4 the following discrete points for SDS are
considered:
• SDS = 0.25g, 0.3g, and 0.35g for the FEE level.
• SDS = 0.4g, 0.5g, 0.6g, 0.8g, 1.0g, 1.25g, 1.5g, and 1.66g for the SEE
level.
Ss=0.60g, SEE(2%/50years)
0.00.10.20.30.40.50.60.70.80.91.0
0 1 2 3 4
SD_6ASD_6BSD_6CSD_6DSD_6E
Periods T (sec)
Site Class A B C D E
SD1-SEE
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Figure 3-3: Design spectral response acceleration map short period – SDS–FEE for site class B.
Figure 3-4: Design spectral response acceleration map short period – SDS-SEE for site class B.
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A family of curves for Soil Site Class A thru E referenced to the short period mapped design spectral acceleration SDS =1.0g is shown in Figure 3-5. The
curves were developed using both the short period and the one-second period
maps.
Ss=1.00g, SEE(2%/50years)
0.0
0.2
0.4
0.6
0.8
1.0
1.2
0 1 2 3 4
SD_4ASD_4BSD_4CSD_4DSD_4E
Periods T (sec)
Site Class A B C D E
Figure 3-5: Design spectra for site class A, B, C, D and E, 5% damping.
The Seismic Performance Category (SPC) definition in the SCDOT
Specifications differs from the AASHTO Division 1-A as follows:
1. The Seismic Performance Category (SPC) is based on the one-second
spectral acceleration at the SEE level Earthquake having a 2%
probability of exceedance in 50 years.
2. The Importance Classification (IC) in the SCDOT Specifications include
three categories of bridges, Critical, Essential, and Normal associated
with IC I, II, and III respectively while AASHTO Division 1-A has two
classifications, IC “I” for Essential bridges and IC “II” for other bridges.
The Specifications are for the design and construction of new bridges to resist
the effects of earthquake motions. The provisions apply to bridges of
conventional slab, beam girder and box girder superstructure construction with
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spans not exceeding 500 ft (150 m). For other types of construction (suspension
bridges, cable-stayed bridges, arch type and movable bridges) and spans
exceeding 500 ft, the SCDOT shall specify and/or approve appropriate
provisions.
Seismic effects for box culverts and buried structures need not be considered,
except when they are subject to unstable ground conditions (e.g., liquefaction,
landslides, and fault displacements) or large ground deformations (e.g., in very
soft ground).
The provisions specified in the specifications are minimum requirements.
Additional provisions are needed to achieve higher performance criteria for
essential or critical bridges. Those provisions are site/project specific and are
tailored based on structure type.
No detailed seismic analysis is required for any single span bridge or for any
bridge in Seismic Performance Category A. For both single span bridges and
bridges classified as SPC A the connections must be designed for specified
forces and must also meet minimum support length requirements.
For SPC B, the displacement demand is checked implicitly against the capacity
without performing an elaborate pushover analysis to determine the
displacement capacity.
For SPC B the displacement capacity, cΔ , is easily obtained for each column
using the following expression:
X( ) 5.3 (.0013)100cHftΔ = ∗ ∗
where,
X DH= Λ
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Λ is a fixity factor for the column equal to: a. Λ = 1 for fixed-free (pinned on one end). b. Λ = 2 for fixed top and bottom.
D = Column Diameter (ft.). H = Height from top of footing to C.G. of superstructure (ft.).
In summary, the objective in developing the new SCDOT Seismic Design
Specification is to balance the required numerical computations to the severity
of the seismic hazard established in SPC A, B, C and D.
Figure 3-6: SPC B Drift Criteria (SCDOT)
Range of Applicability of “Seismic Demand Analysis” Seismic Analysis is conducted in regions where PGA > 9%g following AASHTO
Division 1-A. For illustration of difference in the extent of regions requiring
“Seismic Demand Analysis” following AASHTO Division 1-A and the
recommended specifications, a comparison is performed on the area
surrounding the New Madrid fault and South Carolina. These two areas are
% c
HΔ
DH
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considered since they represent the largest increase in seismic demands when
considering larger return period up to the proposed return period of 2500 years
adopted in NCHRP 12-49. The reference to the increase in seismic demands is
made in relation to AASHTO Division 1-A State of the Practice. Figure 3-7
shows the AASHTO region of required seismic analysis.
Figure 3-7: AASHTO Region of Required Seismic Analysis PGA > 9%
With the selection of the one-second spectral design acceleration spectrum
1a DSS − , the regions of required “Seismic Demand Analysis” vary depending on
the site class (i.e., type of soil) as established in NEHRP 1997 and adopted in
the NCHRP 12-49 document.
Considering a Site Class B for the New Madrid/South Carolina area, the
contour shown in Figure 3-8 in bold black establishes the region of required
“Seismic Demand Analysis” corresponding to the proposed target design
hazard. Based on preliminary selection, the target design hazard is calibrated
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at 2/3 of the spectrum established based on the 2002 USGS hazard maps for a
probability of 2% exceedance in 50 years. The proposed region for Site Class B
of required “Seismic Demand Analysis” is substantially smaller than the
corresponding AASHTO Division 1-A region.
For comparison, the region of required “Seismic Demand Analysis” for Site
Class D is shown in Figure 3-9 for the same area. The proposed region for Site
Class D shows relatively small reduction to the corresponding AASHTO
Division 1-A region.
Figure 3-8: Region of Required “Seismic Demand Analysis” for the Target Design Hazard, Site Class B
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Figure 3-9: Region of Required “Seismic Demand Analysis” for the Target Design Hazard, Site Class D
In comparing the proposed Guidelines to current requirements in AASHTO
Division 1-A, the proposed guidelines fulfill the objective of maintaining to
reducing the number of bridges subjected to “Seismic Demand Analysis.”
Range of Applicability of “Seismic Capacity Analysis” “Seismic Capacity Analysis” is performed for SPC B, C, and D. This analysis is
incremental as follows:
1. SPC B
Implicit displacement capacity check is required similar to SCDOT
Specifications. No Capacity Design is required. This category is
associated with small displacement demand and drifts. Given the
relatively small demands and based on a minimum level of detailing
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identified as Tier II, the bridge structure is expected to perform well,
with its members targeted to remain essentially elastic at ductility level
less than two.
2. SPC C
Implicit displacement capacity deck is required similar to SPC B;
however, setting the acceptance capacity criteria to a higher level of
ductility relative to SPC B. Given the moderate displacement demands
on the bridge structure, a Capacity Design analysis is required in order
to ensure adequate force distribution and proper design for hinging
mechanism. Considering the moderate acceptance criteria, an
incremental Tier III level of detailing is required. An elaborate
pushover analysis is not warranted.
3. SPC D
A pushover analysis is required for this category as a high level of
ductility is expected. Proper distribution of forces and Capacity design
requirements need to be satisfied to ensure a reliable comparison of the
structure displacement capacity against the displacement demands. A
Tier IV level of detailing is required in SPC D.
The contours presented in Section 3.3.1 coincide with SPC B contours. The
same area identified in Section 3.3.1 is used to show the region of required
pushover analysis. By illustrating the region of minimum “Seismic Capacity
Analysis” associated with SPC B and the region of maximum “Seismic Capacity
Analysis” associated with SPC D, the reader can appreciate the incremental
approach proposed for the Guidelines.
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Figures 3-10 and 3-11 shows the Region of required Maximum “Seismic
Capacity Analysis” for the target design hazard for Site Class B and Site Class
D, respectively. A pushover analysis is required in this region.
Figure 3-10: Region of Required Maximum “Seismic Capacity
Analysis” for the Target Design Hazard, Site Class B
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Figure 3-11: Region of Required Maximum “Seismic Capacity
Analysis” for the Target Design Hazard, Site Class D
As shown in Figures 3-10 and 3-11, the region where a pushover analysis is
required is chosen very selectively and is tuned to displacement demands on
the bridge structure. The proposed guidelines aim at fulfilling Task F3-4
objective No. 2 identifying range of applicability for NO Analysis or Limited
Analysis. This approach is a by-product of the steps taken in the NCHRP 12-
49 proposed guidelines and the SCDOT Specifications combined with practical
applications developed and gained in the seismic design practice over the last
decade.
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TASK 4
4 SELECT THE MOST APPROPRIATE DESIGN PROCEDURE FOR STEEL
4.1 General The objective of this task is to select the most appropriate design procedure
(i.e., displacement or force based) for a bridge with a steel superstructure and
to examine both the NCHRP 12-49 and SCDOT using a trial design.
This task emphasis is to address analysis and design requirements for a bridge
with steel girders. The seismic design of a bridge system and components
needs to encompass two categories:
a. System with a restrained connection between the superstructure and the
substructure.
b. System with an unrestrained connection between the superstructure
and the substructure.
Emphasis on the load path and design of various components must be
established recognizing that a lack of consensus may still be present on some
issues.
The 2nd Edition of the LRFD Specifications included for the first time a new
section about the seismic lateral load distribution that discusses the seismic
load path. The focus for these criteria is steel bridges since they normally do
not have monolithic connections as the structural concrete box girder bridges.
1325 NCHRP 20-7(193) Task 6 Report.doc 4-2
The specifications require that a clear and a straightforward load path from
the superstructure to the substructure should exist. All elements that lie in
the load path are primary seismic members and should be designed to stay
elastic during severe ground motions. Diaphragms and cross-frames, lateral
bracing and bearings should be part of the seismic load path. The
specifications suggest that if these members were designed to respond in a
ductile manner or allow some movements, the damage will be limited.
However, the specifications require that the cross frames and end diaphragms
to stay elastic during earthquakes.
On the contrary, NCHRP 12-49 and SCDOT seismic specifications allow for
ductility (i.e., inelastic action) in the superstructure. None of the specifications
contains a uniform and a complete list of primary members identification for
the seismic load path.
4.2 Design Examples Two design examples were selected from the work done by Itani and Sedarat in
2000 entitled “Seismic Analysis and Design of the AISI LRFD Design
Examples of Steel Highway Bridges”. This effort was a continuation to the
1996 AISI published Vol. II Chapter 1B of the Highway Structures, Design
Handbook, “Four LRFD Design Examples of Steel Highway Bridges.” In 1996
these design examples covered the gravity design of the superstructure
according to the AASHTO LRFD Bridge Specifications. The main two
purposes in examining this report is to:
1. Identify the performance objective for seismic design of steel girder
structures.
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2. Identify the specifications utilized for proper completion of the design
process.
Appendix 4A includes the portion of AISI-LRFD report used in this task. This
appendix contains the design calculations as well as the drawing showing
details of each of the two bridges.
Example 1 is a Simple-Span Composite I Girder. The design process shown in
the report includes:
1. Calculation of lateral load at the end cross-frame.
2. The design of the top strut.
3. The design of the diagonal member.
4. The design of the bottom strut.
Two important aspects of the design process are identified:
a. The end cross-frame is designed for the full seismic force with no
reduction of this force assuming a restrained condition of the bridge (i.e.,
shear keys capable of sustaining the full seismic force).
b. A single angle bracing is used for the diagonal member of the end-cross-
frame. As this practice is typical and favored for ease of construction,
the design process for a single angle bracing needs to be referenced or
included for clarity of use by the bridge engineer. AISC has a stand
alone document on “LRFD Design Specification for Single-Angle
Members” that can be included or referenced in the Specifications. This
document is attached in Appendix 4B.
Example 2 is a Two-Span Continuous Composite I Girder. The design process
shown in the report includes:
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1. Calculation of the lateral load at the bent cross-frame.
2. The design of the plate girder connections to the R/C Deck.
3. Design of the top strut.
4. Design of the diagonal member.
5. Design of the bottom strut.
6. Calculation of superstructure lateral capacity.
Three important aspects of the design process are identified:
a. The bent cross-frame is designed to ensure column hinging mechanism
assuming a restrained condition of the superstructure to the bent.
b. The load path from the deck to the girders or the top strut is checked.
c. Double angles with stitches are used for the top strut and the diagonal
member due to the higher seismic demand on this bridge located in
seismic zone 4. AISC LRFD Specifications Chapter E applies to compact
and non-compact prismatic members subject to axial compression
through the centroidal axis. The design process for members with
stitches is also included. The inclusion or reference of the specifications
is needed for clarity and consistency of use by the bridge engineer.
4.3 Load Path and Performance Criteria Specifications regarding the load path for a slab-on-girder bridge are examined
using SCDOT and NCHRP 12-49 documents. SCDOT specifications has a
general section on load path while NCHRP 12-49 has a section only on “Ductile
End-Diaphragm in Slab-on-Girder Bridge.” The section from SCDOT
1325 NCHRP 20-7(193) Task 6 Report.doc 4-5
specifications is included in Appendix 4D while the section from NCHRP 12-49
is included in Appendix 4E.
As seen from examining both of these documents, it is important to
differentiate between ordinary bracing referred to in the SCDOT specifications
and specially detailed diaphragms referred to in the NCHRP 12-49. The AISC
provisions limit the force reduction factor R to 3 for ordinary bracing that is a
part of a seismic resisting system not satisfying the special seismic provisions.
It is proposed to adopt the AISC limit for an R reduction factor of 3. Special
end-diaphragm addressed in NCHRP 12-49 will be considered for bracing
system with a reduction factor, R, greater than 3 as stipulated in the AISC
provisions.
Section 7.1 and 7.2 of SCDOT specifications will be enhanced for general
treatment of load path and Performance Criteria. The following is a
duplication of these two sections.
General
“The Engineer shall demonstrate that a clear, straight-forward load
path to the substructure exists and that all components and connections
are capable of resisting the imposed seismic load effects consistent with
the chosen load path.
The flow of forces (see Figure 4.1) in the assumed load path must be
accommodated through all affected components and details including,
but not limited to, flanges and webs of main beams or girders, cross-
frames, steel-to-steel connections, slab-to-steel interfaces, and all
components of the bearing assembly from bottom flange interface
through the confinement of anchor bolts or similar devices in the
1325 NCHRP 20-7(193) Task 6 Report.doc 4-6
substructure. The substructure shall also be designed to transmit the
imposed force effects into the ground.
a) Pile Footing b) Drilled Shaft
Figure 4-1: Seismic Load Path and Affected Components
The design of end diaphragms and cross-frames shall include analysis
cases with horizontal supports at an appropriate number of bearings,
consistent with Section 7.7.2 of SCDOT Specifications.
A viable load path shall be established to transmit the inertial loads to
the foundation based on the stiffness characteristics of the deck,
diaphragms, cross-frames, and lateral bracing. Unless a more refined
analysis is made, an approximate load path shall be assumed as follows:
The following requirements apply to bridges with either: Formatted: Bullets and Numbering
1325 NCHRP 20-7(193) Task 6 Report.doc 4-7
• a concrete deck that can provide horizontal diaphragm action or
• a horizontal bracing system in the plane of the top flange.
The seismic loads in the deck shall be assumed to be transmitted
directly to the bearings through end diaphragms or cross-frames. The
development and analysis of the load path through the deck or through
the top lateral bracing, if present, shall utilize assumed structural
actions analogous to those used for the analysis of wind loadings.”
Criteria
“This section is intended for design of superstructure steel components.
Those components are classified into two categories: Ductile and
Essentially Elastic. Based on the characteristics of the bridge structure,
the designer has one of three choices:
• Type 1 – Design a ductile substructure with an essentially elastic
superstructure.
• Type 2 – Design an essentially elastic substructure with a ductile
superstructure.
• Type 3 – Design an elastic superstructure and substructure with
a fusing mechanism at the interface between the superstructure
and the substructure.
For Type 1 choice, the designer shall refer back to Section 8 of this
document on designing for a ductile substructure. For Type 2 choice, the
design of the superstructure is accomplished using a force reduction
approach. Those factors are used for the design of transverse bracing
members, top laterals and bottom laterals. The reduction factors shown
in Table 7.1 shall be used.
1325 NCHRP 20-7(193) Task 6 Report.doc 4-8
Table 4-1: Reduction Factors for Steel Superstructure Bracings
Essential or
Critical Bridges Normal Bridges
Functional Evaluation 1 2 Safety Evaluation 2 4
For Type 3 choice, the designer shall assess the overstrength capacity for
the fusing interface including shear keys and bearings, then design for
an essentially elastic superstructure and substructure. The minimum
overstrength lateral design force shall be calculated using an
acceleration of 0.4 g or the elastic seismic force whichever is smaller. If
isolation devices are used, the superstructure shall be designed as
essentially elastic (see Section 7.6 of SCDOT Specifications).
In this section, reference to an essentially elastic component is used
where the force demand to capacity ratio of any member in the
superstructure is less than 1.3.”
4.4 Summary In reviewing the SCDOT specifications, the NCHRP 12-49, and the AISI LRFD
examples, the following recommendations are proposed:
1. Adopt AISC LRFD Specifications for design of single angle members and
members with stitches.
2. Allow for three types of a bridge structural system as adopted in SCDOT
Specifications.
3. Adopt a force reduction factor of 3 for design of normal end cross-frame.
3
1325 NCHRP 20-7(193) Task 6 Report.doc 4-9
4. Adopt NCHRP 12-49 for design of “Ductile End-Diaphragm” where a
force reduction factor greater than 3 is desired.
1325 NCHRP 20-7(193) Task 6 Report.doc 4A-1
TASK 4
APPENDIX 4A
CENTER FOR CIVIL ENGINEERING AND EARTHQUAKE RESEARCH
Report No. CCEER 00-8
Seismic Analysis and Design of the
AISI LRFD Design Examples
of Steel Highway Bridges
Ahmad M. Itani
Hassan Sedarat
• Reno
Engineering Research and Development Center
College of Engineering
University of Nevada, Reno
1325 NCHRP 20-7(193) Task 6 Report.doc 4B-1
TASK 4
APPENDIX 4B
LOAD AND RESISTANCE FACTOR DESIGN SPECIFICATION FOR
SINGLE-ANGLE MEMBERS
1325 NCHRP 20-7(193) Task 6 Report.doc 4C-1
TASK 4
APPENDIX 4C
CHAPTER E – COLUMNS AND OTHER COMPRESSION MEMBERS
1325 NCHRP 20-7(193) Task 6 Report.doc 4D-1
TASK 4
APPENDIX 4D
SOUTH CAROLINA DEPARTMENT OF TRANSPORTATION
SPECIFICATIONS FOR HIGHWAY BRIDGES
1325 NCHRP 20-7(193) Task 6 Report.doc 4E-1
TASK 4
APPENDIX 4E
NCHRP 12-49
1325 NCHRP 20-7(193) Task 6 Report.doc 4E-2
7.7 Structural Steel Design Requirements
7.7.8.2 Ductile End-Diaphragm in Slab-on-Girder Bridge
Ductile end-diaphragms in slab-on-girder bridges can be designed to be the
ductile energy dissipating elements for seismic excitations in the transverse
directions of straight bridges provided that:
a. specially detailed diaphragms capable of dissipating energy in a stable
manner and without strength degradation upon repeated cyclic testing
are used;
b. only ductile energy dissipating systems whose adequate seismic
performance has been proven through cycling inelastic testing are used;
c. the design considers the combined and relative stiffness and strength of
end-diaphragms and girders (together with their bearing stiffeners) in
establishing the diaphragms strength and design forces to consider for
the capacity protected elements;
d. the response modification factor to be considered in design of the ductile
diaphragm is given by:
1
DED
SUB
DED
SUB
KKR KK
μ⎛ ⎞+⎜ ⎟⎜ ⎟=⎜ ⎟+⎜ ⎟⎝ ⎠
(7.7.8.2-1)
where μ is the ductility capacity of the end-diaphragm itself, and
KDED/KSUB is the ratio of the stiffness of the ductile end-diaphragms and
substructure (unless the designer can demonstrate otherwise, μ should
not be taken greater than 4);
e. all details/connections of the ductile end-diaphragms are welded;
1325 NCHRP 20-7(193) Task 6 Report.doc 4E-3
f. the bridge does not have horizontal wind-bracing connecting the bottom
flanges of girders, unless the last wind bracing panel before each support
is designed as a ductile panel equivalent and in parallel to its adjacent
vertical end-diaphragm; and
g. an effective mechanism is present to ensure transfer of the inertia-
induced transverse horizontal seismic forces from the slab to the
diaphragm.
Overstrength factors to be used to design the Capacity Protected Elements
depend on the type of ductile diaphragm used, and shall be based on available
experimental research results.
8.7 Structural Steel Design Requirements
8.7.8.2 Ductile End-Diaphragm in Slab-on-Girder Bridge
Ductile end-diaphragms in slab-on-girder bridges can be designed to be the
ductile energy dissipating elements for seismic excitations in the transverse
directions of straight bridges provided that:
a. Specially detailed diaphragms capable of dissipating energy in a stable
manner and without strength degradation upon repeated cyclic testing
are used;
b. Only ductile energy dissipating systems whose adequate seismic
performance has been proven through cycling inelastic testing are used;
c. Design considers the combined and relative stiffness and strength of
end-diaphragms and girders (together with their bearing stiffeners) in
establishing the diaphragms strength and design forces to consider for
the capacity protected elements;
1325 NCHRP 20-7(193) Task 6 Report.doc 4E-4
d. The response modification factor to be considered in design of the ductile
diaphragm is given by:
1
DED
SUB
DED
SUB
KKRKK
μ⎛ ⎞+⎜ ⎟⎜ ⎟=⎜ ⎟+⎜ ⎟⎝ ⎠
(8.7.8.2-1)
where μ is the ductility capacity of the end-diaphragm itself, and
KDED/KSUB is the ratio of the stiffness of the ductile end-diaphragms and
substructure; unless the engineer can demonstrated otherwise, μ should
not be taken greater than 4;
e. All details/connections of the ductile end-diaphragms are welded.
f. The bridge does not have horizontal wind-bracing connecting the bottom
flanges of girders, unless the last wind bracing panel before each support
is designed as a ductile panel equivalent and in parallel to its adjacent
vertical end-diaphragm.
g. An effective mechanism is present to ensure transfer of the inertia-
induced transverse horizontal seismic forces from the slab to the
diaphragm.
Overstrength factors to be used to design the capacity-protected elements
depend on the type of ductile diaphragm used, and shall be based on available
experimental research results.
1325 NCHRP 20-7(193) Task 6 Report.doc 5-1
TASK 5
5 RECOMMEND LIQUEFACTION DESIGN PROCEDURE
5.1 Objective The objective of this task is to review applicable recent research and
information currently available on liquefaction and to recommend design
procedures consistent with the “Displacement Approach” adopted for the
proposed specifications. The proposed approach is to streamline the provisions
provided by NCHRP 12-49 in one separate section or appendix. The extent of
the provisions are established in light of the overall methodology and
framework established in the tasks:
a. Task 2 – Finalize Seismic Hazard Level
b. Task 3 – Expand the Extent of the “No Analysis” Zone
5.2 NCHRP 12-49 Liquefaction Design Requirements
NCHRP 12-49 added considerable amount of information for the provisions on
liquefaction. The general design approach outlined in NCHRP 12-49 consists
of the following:
1. Specific design requirements for piled foundations, drilled shafts and
spread footing exposed to liquefaction with no lateral flow.
1325 NCHRP 20-7(193) Task 6 Report.doc 5-2
2. For the above mentioned types of foundations subjected to lateral flow,
proceed with the following steps:
a. Design the piles or spread footings to resist the forces generated
by the lateral spreading.
b. If the structure cannot be designed to resist the forces, assess
whether the structure is able to tolerate the anticipated
movements and meet the geometric and structural constraints of
the provisions.
c. If the structure cannot meet the performance requirements of the
provisions, assess the costs and benefits of various mitigation
measures to minimize the movements to a tolerable level to meet
the desired performance objective.
Appendix 5A contains NCHRP 12-49 requirements for Foundation Design and
Liquefaction Design for SDR 3 (Chapter 7 of NCHRP 12-49) and SDR 4
(Chapter 8 of NCHRP 12-49).
In adopting a “Displacement” Approach for the new specifications and
considering a No Collapse Criteria, the new specifications will be altered in
determining the adequacy of the structure based strictly on the displacement
demands. Minimum strength requirements would be introduced to minimize
the effects of any geometric non-linearities. Provisions related to steps a) and
c) mentioned above and related to a “Force Based Approach” will be eliminated
for consistency with the overall approach.
1325 NCHRP 20-7(193) Task 6 Report.doc 5-3
5.3 Damage Severity in Past Earthquakes In order to gain insight on the damage severity on bridges during past
earthquakes, the catalog on the seismic performance of bridges in the presence
of liquefaction-induced ground displacement authored by Stephen A.
Dickenson, Nason J. McCullough, Mark G. Barkau, and Bryan J. Wavra is
used.
Each bridge in this catalog has been assigned a damage severity rating DSR
according to the classification scheme outlined in Table 5-1. A summary of this
catalog is shown in Table 5-2.
Table 5-1: Damage Severity Description
DAMAGE SEVERITY RATING (DSR) DAMAGE DESCRIPTION
DSR = 3 Severe Damage: Abutments moved streamward and/or markedly subsided;
piers shifted, tilted, settled, or fell over. Large movements of foundation units. Substructure rendered unsalvable.
DSR = 2 Moderate Damage: Distinct and measurable net displacements as in
previous category but to a lesser degree, so that the substructure could perhaps be repaired and used to support a new superstructure.
DSR = 1 Minor Damage: Evidence of foundation movements such as cracked
backwalls, split piles, and closed expansion devices, but net displacements small and substructure serviceable. Minor abutment slumping.
DSR = 0 Nil Damage: No evidence of foundation displacements.
1325 NCHRP 20-7(193) Task 6 Report.doc 5-4
Table 5-2: Damage Severity Rating vs. Earthquake Magnitude
Earthquake Mw DSR Minimum DSR Maximum1995 Manzanillo, Mexico 7.5 1 11995 Hyogo-ken-Nanbu (Kobe), Japan 6.9 0 31994 Northridge 6.7 0 01994 Mindoro Island, Phillipines 7.1 3 31993 Island of Guam 8.4 1 11993 Hokkaido Nansei-oki, Japan 7.8 0 21992 Erzincan, Turkey 6.7 1 21991 Costa Rica 7.4 0 31990 Luzon, Phillipines 7.9 1 31989 Loma Prieta 6.9 0 11983 Nihonkai-Chuba 7.7 0 11980 El-Asnam, Algeria 7.2 1 11979 Imperial Valley, California 6.5 1 11978 Miyagi-Ken-oki, Japan 7.3 1 21976 Mindanao, Phillipines 7.9 1 11976 Tangshan, China 7.8 3 31975 Haicheng, China 7.2 3 31968 Ebino 6.1 1 11964 Alaska 9.21964 Niigata, Japan 7.3 3 31948 Fukui, Japan 6.9 2 31923 Kanto, Japan 6.9 2 31906 San Francisco 7.9 0 31886 Charleston, South Carolina ? 3 3
The full catalog is included in Appendix 5B. As seen from Table 5-2 a DSR
equal to 2 corresponding to moderate damage is associated with an earthquake
magnitude Mw of 6.7 or higher while a DSR equal to 3 corresponding to severe
damage is associated with an earthquake magnitude Mw of 6.9 or higher.
5.4 Proposed Liquefaction Design Requirements An evaluation of the potential for and consequences of liquefaction within near
surface soil shall be made in accordance with the following requirements:
Liquefaction is required for a bridge in SPC D unless one of the following
conditions is met:
1325 NCHRP 20-7(193) Task 6 Report.doc 5-5
a. The mean magnitude for the 5% PE in 50-year event is less than 6.5.
b. The mean magnitude for the 5% PE in 50-year event is less than 6.7
and the normalized Standard Penetration Test (SPT) blow count
[(N1)60] is greater than 20.
Procedures given in Appendix D of NCHRP 12-49 and adopted from California
DMG Special Publication 117 shall be used to evaluate the potential for
liquefaction.
If it is determined that liquefaction can occur at a bridge site then the bridge
shall be supported on deep foundations or the ground improved so that
liquefaction does not occur. If liquefaction occurs then the bridge shall be
designed and analyzed in two configurations as follows:
1. Nonliquefied Configuration: The structure shall be analyzed and
designed, assuming no liquefaction occurs using the ground response
spectrum appropriate for the site soil conditions.
2. Liquefaction Configuration: The structure as designed in Nonliquefied
Configuration above shall be reanalyzed and redesigned, if necessary,
assuming that the layer has liquefied and the liquefied soil provides
whatever residual resistance is appropriate (i.e., “p-y curves” or modulus
of sub-grade reaction values for lateral pile response analyses consistent
with liquefied soil conditions). The design spectra shall be the same as
that used in Nonliquefied Configuration unless a site-specific response
spectra has been developed using nonlinear, effective stress methods
(e.g., computer program DESRA or equivalent) that properly account for
the buildup in pore-water pressure and stiffness degradation in
liquefiable layers. The reduced response spectra resulting from the site-
1325 NCHRP 20-7(193) Task 6 Report.doc 5-6
specific nonlinear, effective stress analyses shall not be less than 2/3’s of
that used in Nonliquefied Configuration.
The Designer shall cover explicit detailing of plastic hinging zones for both
cases mentioned above since it is likely that locations of plastic hinges for the
Liquefied Configuration are different than locations of plastic hinges for the
Non-Liquefied Configuration. Design requirements of SPC “D” including shear
reinforcement shall be met for the Liquefied and Non-Liquefied Configuration.
5.5 Summary The following list highlights the main proposed liquefaction design
requirements:
a. Liquefaction design requirements are applicable to SPC “D”.
b. Liquefaction design requirements are dependent on the mean magnitude
for the 5% PE in 50-year event and the normalized Standard
Penetration Test (SPT) blow count [(N1)60].
c. If liquefaction occurs, then the bridge shall be designed and analyzed for
the Liquefied and Non-Liquefied configurations.
Design requirements for lateral flow are still debatable and have not reached a
consensus worth comfortably adopting. The IAI geotechnical team is preparing
a task to address this topic and complement the effort produced in the NCHRP
12-49 document.
1325 NCHRP 20-7(193) Task 6 Report.doc 5A-1
TASK 5
APPENDIX 5A
NCHRP 12-49
PART I: SPECIFICATIONS 2003 GUIDELINES FOR THE SEISMIC DESIGN OF HIGHWAY BRIDGES
SECTION 7 72 MCEER/ATC-49
7.4 FOUNDATION DESIGN REQUIREMENTS
7.4.1 Foundation Investigation
7.4.1.1 General
A subsurface investigation, including borings and laboratory soil tests, shall be conducted in accordance with the provisions of Appendix B to provide pertinent and sufficient information for the determination of the Site Class of Article 3.4.2.1. The type and cost of foundations should be considered in the economic, environmental, and aesthetic studies for location and bridge type selection.
7.4.1.2 Subsurface Investigation
Subsurface explorations shall be made at pier and abutment locations, sufficient in number and depth, to establish a reliable longitudinal and transverse substrata profile. Samples of material encountered shall be taken and preserved for future reference and/or testing. Boring logs shall be prepared in detail sufficient to locate material strata, results of penetration tests, groundwater, any artesian action, and where samples were taken. Special attention shall be paid to the detection of narrow, soft seams that may be located at stratum boundaries.
7.4.1.3 Laboratory Testing
Laboratory tests shall be performed to determine the strength, deformation, and flow characteristics of soils and/or rocks and their suitability for the foundation selected. In areas of higher seismicity (e.g., SDR 3, 4, 5, and 6), it may be appropriate to conduct special dynamic or cyclic tests to establish the liquefaction potential or stiffness and material
damping properties of the soil at some sites, if unusual soils exist or if the foundation is supporting a critical bridge.
7.4.2 Spread Footings
Spread footing foundations for SDR 3 shall be designed using column loads developed by capacity design principles or elastic seismic loads, in accordance with Strength Limit State requirements given in Article 10.6.3 of the LRFD Bridge Design Specifications (AASHTO, 1998a, and subsequent amendments), hereinafter referred to as the AASHTO LRFD provisions. It will not normally be necessary to define spring constants for displacement evaluations or moment-rotation and horizontal force-displacement behavior of the footing-soil system (Article 5.3.4). Checks shall also be made to confirm that flow slides and loss of bearing support from liquefaction do not occur (Article 7.6).
7.4.2.1 Moment and Shear Capacity
The overturning capacity of the spread footings shall be evaluated using 1.0 times the nominal moment capacity of the column (Article 4.8) or the elastic seismic design force within the column, whichever is less. Procedures for Strength Limit State Design given in Article 10.6.3 of the AASHTO LRFD provisions shall be used when performing this evaluation.
A triangular elastic stress distribution within the soil shall be used. The peak bearing soil pressure for the triangular distribution shall not exceed the ultimate bearing capacity of the soil at the toe of the footing. The width of maximum liftoff shall be no greater than 1/2 of the footing width for moment loading in each of the two directions treated separately.
PART I: SPECIFICATIONS 2003 GUIDELINES FOR THE SEISMIC DESIGN OF HIGHWAY BRIDGES
SECTION 7 73 MCEER/ATC-49
If a non-triangular stress distribution occurs or if the liftoff is greater 1/2 of the footing, either the footing shall be re-sized to meet the above criteria or special studies shall be conducted to demonstrate that non-triangular stress pressure distribution or larger amounts of liftoff will not result in excessive permanent settlement during seismic loading. The special studies shall include push-over analyses with nonlinear foundation springs for SDAP E conditions.
No shear capacity evaluation of the footing will normally be required for SDR 3.
7.4.2.2 Liquefaction Check
An evaluation of the potential for liquefaction within near-surface soil shall be made in accordance with requirements given in Article 7.6 and Appendix D of these Specifications. If liquefaction is predicted to occur for the design earthquake, the following additional requirements shall be satisfied: Liquefaction without Lateral Flow or Spreading
For sites that liquefy but do not undergo lateral flow or spreading, the bottom of the spread footing shall be located either below the liquefiable layer or at least twice the minimum foundation width above the liquefiable layer. If liquefaction occurs below the footing, settlements resulting from the dissipation of excess porewater pressures shall be established in accordance with procedures given in Appendix D.
If the depth of the liquefiable layer is less than twice the minimum foundation width, spread footing foundations shall not be used, unless • ground improvement is performed to mitigate
the occurrence of liquefaction, or • special studies are conducted to demonstrate
that the occurrence of liquefaction will not be
detrimental to the performance of the bridge support system.
Before initiating any evaluations of ground improvement alternatives or before conducting special studies, the potential applicability of deep foundations as an alternative to spread footings shall be discussed with the owner. Liquefaction with Lateral Flow or Spreading
If lateral flow or lateral spreading is predicted to occur, the amount of displacement associated with lateral flow or lateral spreading shall be established in accordance with procedures given in Appendix D. Once the deformation has been quantified, the following design approach shall be used. • Determine whether the spread footings can be
designed to resist the forces generated by the lateral spreading without unusual size or design requirements.
• If the footing cannot resist forces from lateral spreading or flow, assess whether the structure is able to tolerate the anticipated movements and meet the geometric and structural constraints of Table C3.2-1. The maximum plastic rotation shall be as defined in Article 7.7.9 and 7.8.6.
• If the structure cannot meet the performance requirements of Table 3.2-1, assess the costs and benefits of various mitigation measures to minimize the movements to a level that will meet the desired performance objective. If a higher performance is desired so that the spread footings will not have to be replaced, the allowable plastic rotations for concrete columns given in Article 7.7.9 and 7.8.6 shall be met.
The owner shall be apprised of and concur with the approach used for the design of spread footing foundations for lateral flow or lateral spreading conditions.
PART I: SPECIFICATIONS 2003 GUIDELINES FOR THE SEISMIC DESIGN OF HIGHWAY BRIDGES
SECTION 7 74 MCEER/ATC-49
7.4.3 Driven Piles
7.4.3.1 General
Resistance factors for pile capacities shall be as specified in Table 10.5.4-2 of the AASHTO LRFD provisions, with the exception that resistance factors of 1.0 shall be used for seismic loads.
For the effect of settling ground and downdrag loads, unfactored load and resistance factors (γ = 1.0; φ = 1.0) shall be used, unless required otherwise by the owner.
Batter piles shall not be used where downdrag loads are expected unless special studies are performed.
For seismic loading the groundwater table location shall be the average groundwater location, unless the owner approves otherwise.
7.4.3.2 Design Requirements
Driven pile foundations subject to SDR 3 shall be designed for column moments and shears developed in accordance with the principles of capacity design (Article 4.8) or the elastic design forces, whichever is smaller. The Strength Limit State requirements given in Article 10.7.3 of the AASHTO LRFD provisions shall apply for design.
With the exception of pile bents, it will not normally be necessary to define spring constants for displacement evaluations or moment-rotation and horizontal force-displacement analyses for SDR 3 (Article 5.3.4). For pile bents, the estimated depth of fixity shall be used in evaluating response.
If liquefaction is predicted at the site, the potential effects of liquefaction on the capacity of the driven pile foundation system
shall be evaluated in accordance with procedures given in Article 7.4.3.4.
7.4.3.3 Moment and Shear Design
The capacity of the geotechnical elements of driven pile foundations shall be designed using 1.0 times the nominal moment capacity of the column or the elastic design force within the column (Article 4.8), whichever is smaller. Unfactored resistance (φ = 1.0) shall be used in performing the geotechnical capacity check. The loads on the leading row of piles during overturning shall not exceed the plunging capacity of the piles. Separation between the pile tip and the soil (i.e. gapping) shall be allowed only in the most distant row of piles in the direction of loading. Forces on all other rows of piles shall either be compressive or not exceed the nominal tension capacity of the piles.
If the plunging capacity of the leading pile is exceeded or if uplift of other than the trailing rows of piles occurs (see Figure C3.3.1-2), special studies shall be conducted to show that performance of the pile system is acceptable. These studies shall be performed only with the prior consent of the owner and SDAP E is required.
Structural elements of pile foundations shall be designed using the overstrength moment capacity of the column or the elastic design force within the column (Article 4.8), whichever is smaller.
The maximum shear force on the pile(s) shall be less than the structural shear capacity of the piles.
7.4.3.4 Liquefaction Check
An evaluation of the potential for liquefaction shall be made in accordance with requirements given in Article 7.6 and
PART I: SPECIFICATIONS 2003 GUIDELINES FOR THE SEISMIC DESIGN OF HIGHWAY BRIDGES
SECTION 7 75 MCEER/ATC-49
Appendix D of these Specifications. If liquefaction is predicted to occur for the design earthquake, the following additional requirements shall be satisfied: Liquefaction without Lateral Flow or Spreading
• The pile shall penetrate beyond the bottom of the liquefied layer by at least 3 pile diameters or to a depth that axial and lateral pile capacity are not affected by liquefaction of the overlying layer, whichever is deeper.
• The shear reinforcement in a concrete or pre-stressed concrete pile shall meet the requirements of Sec 7.8.2.3 from the pile or bent cap to a depth of 3 diameters below the lowest liquefiable layer.
• Effects of downdrag on the pile settlements shall be determined in accordance with procedures given in Appendix D.
Liquefaction with Lateral Flow or Lateral Spreading
• Design the piles to resist the forces generated by the lateral spreading.
• If the forces cannot be resisted, assess whether the structure is able to tolerate the anticipated movements and meet the geometric and structural constraints of Table C3.2-1. The maximum plastic rotation of the piles shall be as defined in Article 7.7.9 and Article 7.8.6.
• If the structure cannot meet the performance requirements of Table 3.2-1, assess the costs and benefits of various mitigation measures to reduce the movements to a tolerable level to meet the desired performance objective. If a higher performance is desired so that the piles will not have to be replaced, the allowable plastic rotations of Articles 7.7.9.2 and 7.8.6.2 shall be met.
7.4.4 Drilled Shafts
Procedures identified in Article 7.4.3.2, including those for liquefaction and dynamic settlement, shall be applied with the exception that the ultimate capacity in compression or uplift loading for single shaft foundations in SDR 3 shall not be exceeded during maximum seismic loading without special design studies and the owner’s approval. The flexibility of
the drilled shaft shall also be represented in the design using either the estimated depth of fixity or soil springs in a lateral pile analysis.
Diameter adjustments shall be considered during lateral load analyses of shafts with a diameter greater than 600 mm if the shaft is free to rotate, as in the case of a column extension (i.e., no pile cap). Contributions from base shear shall also be considered.
PART I: SPECIFICATIONS 2003 GUIDELINES FOR THE SEISMIC DESIGN OF HIGHWAY BRIDGES
MCEER/ATC-49 79 SECTION 7
7.6 LIQUEFACTION DESIGN REQUIREMENTS
7.6.1 General
An evaluation of the potential for and consequences of liquefaction within near-surface soil shall be made in accordance with the following requirements. A liquefaction assessment is required unless one of the following conditions is met or as directed otherwise by the owner. • Mean magnitude for the Maximum
Considered Earthquake (MCE) is less than 6.0 (Figures 7.6.1-1 to 7.6.1-4);
• Mean magnitude of the MCE is less than 6.4 and equal to or greater than 6.0, and the normalized Standard Penetration Test (SPT) blow count [(N1)60] is greater than 20;
• Mean magnitude for the MCE is less than 6.4 and equal to or greater than 6.0, (N1)60 is greater than 15, and FaSs is between 0.25 and 0.375.
If the mean magnitude shown in Figures 7.6.1-1 to 7.6.1-4 is greater than or equal to 6.4, or if the above requirements are not met for magnitudes between 6.0 and 6.4, or if for the Expected Earthquake, FaSs is greater than 0.375, evaluations of liquefaction and associated phenomena such as lateral flow, lateral spreading, and dynamic settlement shall be evaluated in accordance with these Specifications. 7.6.2 Evaluation of Liquefaction Potential
Procedures given in Appendix D shall be used to evaluate the potential for liquefaction.
7.6.3 Evaluation of the Effects of Liquefaction and Lateral Ground Movement
Procedures given in Appendix D shall be used to evaluate the potential for and effects of liquefaction and liquefaction-related permanent ground movement (i.e., lateral spreading, lateral flow, and dynamic settlement). If both liquefaction and ground movement occur, they shall be treated as separate and independent load cases, unless agreed to or directed otherwise by the owner.
7.6.4 Design Requirements if Liquefaction and Ground Movement Occurs
If it is determined from Appendix D that liquefaction can occur at a bridge site, then one or more of the following approaches shall be implemented in the design.
If liquefaction and no lateral flow occurs, then the bridge shall be designed by conventional procedures including the following requirements: a. Piled Foundations, Drilled Shafts and Pile
Bents: The pile or shaft shall penetrate beyond the bottom of the liquefied layer by at least 3 pile diameters or to a depth that is not affected by liquefaction of the overlying layer or by partial build-up in pore-water pressure, whichever is deeper. In addition the shear reinforcement in a concrete or pre-stressed concrete pile shall meet the requirements of Sec 7.8.2.3 from the pile or bent cap to a depth of 3 diameters below the lowest liquefiable layer.
b. Spread Footings: The bottom of the spread footing shall either be below the liquefiable layer or it shall be at least twice the minimum foundation width of the footing above the liquefiable layer. If liquefaction occurs beneath the base of the footing, the magnitude of settlement caused by liquefaction shall be estimated, and its effects on bridge performance assessed.
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If lateral flow or lateral spreading is predicted to occur, the following options shall be considered as detailed in Appendix D. 1. Design the piles or spread footings to resist the
forces generated by the lateral spreading. 2. If the structure cannot be designed to resist the
forces, assess whether the structure is able to tolerate the anticipated movements and meet the geometric and structural constraints of Table C3.2-1. The maximum plastic rotation
of the piles shall be as defined in Article 7.7.9 and 7.8.6.
3. If the structure cannot meet the performance requirements of Table 3.2-1, assess the costs and benefits of various mitigation measures to minimize the movements to a tolerable level to meet the desired performance objective. If a higher performance is desired so that the spread footings or piles will not have to be replaced, the allowable plastic rotations of Articles 7.7.9.2 and 7.8.6.2 shall be met.
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Figure 7.6.1-1 Mean Earthquake Magnitude Map for Western United States
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Figure 7.6.1-2 Mean Earthquake Magnitude Map for Central and Eastern United States
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Figure 7.6.1-3 Mean Earthquake Magnitude Map for Northwest Alaska
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Figure 7.6.1-4 Mean Earthquake Magnitude Map for Southeast Alaska
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7.6.5 Detailed Foundation Design Requirements
Article 7.4 contains detailed design requirements for each of the different foundation types.
7.6.6 Other Collateral Hazards
The potential occurrence of collateral hazards resulting from fault rupture, landsliding, differential ground compaction, and flooding and inundation shall be evaluated. Procedures for making these evaluations are summarized in Appendix D.
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SECTION 8 104 MCEER/ATC-49
8.4 FOUNDATION DESIGN REQUIREMENTS
8.4.1 Foundation Investigation
8.4.1.1 General
A subsurface investigation, including borings and laboratory soil tests, shall be conducted in accordance with the provisions of Appendix B to provide pertinent and sufficient information for the determination of the Site Class of Article 3.4.2.1. The type and cost of foundations should be considered in the economic, environmental, and aesthetic studies for location and bridge type selection.
8.4.1.2 Subsurface Investigation
Subsurface explorations shall be made at pier and abutment locations, sufficient in number and depth, to establish a reliable longitudinal and transverse substrata profile. Samples of material encountered shall be taken and preserved for future reference and/or testing. Boring logs shall be prepared in detail sufficient to locate material strata, results of penetration tests, groundwater, any artesian action, and where samples were taken. Special attention shall be paid to the detection of narrow, soft seams that may be located at stratum boundaries.
8.4.1.3 Laboratory Testing
Laboratory tests shall be performed to determine the strength, deformation, and flow characteristics of soils and/or rocks and their suitability for the foundation selected. In areas of higher seismicity (e.g., where SDR 4, 5, and 6 apply), it may be appropriate to conduct special dynamic or cyclic tests to establish the liquefaction potential or stiffness and material
damping properties of the soil at some sites, if unusual soils exist or if the foundation is supporting a critical bridge.
8.4.2 Spread Footings
The design of spread footing foundations located in SDR 4, 5, and 6 shall be based on column moments and shears developed using capacity design principles as described in Section 4.8.
Foundation flexibility (Article 5.3.4) shall be modeled for Soil Types C, D, and E if foundation flexibility results in more than a 20% change in response (see Article C5.3.4). For Soil Types A and B, soil flexibility does not need to be considered because of the stiffness of the soil or rock. The potential for and effects of liquefaction and dynamic settlement shall also be determined for spread footing foundations subject to SDR 4 and above. Normally, spread footings shall not be located at SDR 4, 5, and 6 sites where liquefaction is predicted to occur, unless: • the foundation is located below the liquefiable
layer.
• it can be demonstrated by special studies that liquefaction and its effects are very limited, or
• the ground will be improved such that liquefaction will not occur.
Owner approval shall be obtained before proceeding with a spread footing design at a site where liquefaction is predicted to occur.
8.4.2.1 Spring Constants for Footing (Nonliquefiable Sites)
When required to represent foundation flexibility, spring constants shall be developed for spread footing using equations given in Tables 8.4.2.1-1 and 8.4.2.1-2. Alternative procedures given in the FEMA 273 Guidelines for the Seismic Rehabilitation of Buildings
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(ATC/BSSC, 1997) are also suitable for estimating spring constants. These computational methods are appropriate for sites that do not liquefy or lose strength during earthquake loading. See Article 8.4.2.3 for sites that are predicted to liquefy.
The shear modulus (G) used to compute the stiffness values in Table 8.4.2.1-1 shall be determined by adjusting the low-strain shear modulus (Gmax) for the level of shearing strain using the following strain adjustment factors, unless other methods are approved by the owner. For FvS1 ≤ 0.40:
• G/Gmax = 0.50 for Expected Earthquake ground motions
• G/Gmax = 0.25 for Maximum Considered Earthquake (MCE) ground motions
For FvS1 > 0.40: • G/Gmax = 0.25 for Expected Earthquake
ground motions • G/Gmax = 0.10 for MCE ground motions
Uplift shall be allowed for footings subject to SDR 4, 5, and 6. The following area adjustment factors (Ra) shall be applied to the equivalent area to account for geometric
nonlinearity introduced by uplift, unless the Owner approves otherwise. For FvS1 ≤ 0.40: • Ra = 1.0 for Expected Earthquake ground
motions
• Ra = 0.75 for MCE ground motions
For FvS1 > 0.40: • Ra = 0.75 for Expected Earthquake ground
motions
• Ra = 0.5 for MCE ground motions
Values of Gmax shall be determined by seismic methods (e.g., crosshole, downhole, or SASW), by laboratory testing methods (e.g., resonant column with adjustments for time), or by empirical equations (Kramer, 1996). The uncertainty in determination of Gmax shall be considered when establishing strain adjustment factors.
No special computations are required to determine the geometric or radiation damping of the foundation system. Five percent system damping shall be used for design, unless special studies are performed and approved by the owner.
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Table 8.4.2.1-1 Surface Stiffnesses for a Rigid Plate on a Semi-Infinite Homogeneous Elastic
Half-Space (Adapted from Gazetas, 1991)1
Stiffness Parameter Rigid Plate Stiffness at Surface, Ki'
Vertical Translation, Kz'
( )ν
⎡ ⎤+⎢ ⎥− ⎣ ⎦
0.75
0.73 1.541GL B
L
Horizontal Translation, Ky' (toward long side) ( )ν
⎡ ⎤+⎢ ⎥− ⎣ ⎦
0.85
2 2.52GL B
L
Horizontal Translation, Kx' (toward short side) ( )ν ν
⎡ ⎤⎡ ⎤ ⎛ ⎞+ − −⎢ ⎥⎜ ⎟⎢ ⎥− − ⎝ ⎠⎣ ⎦ ⎣ ⎦
0.85
2 2.5 0.1 12 0.75GL GL BB
L L
Rotation, Kθx' (about x axis) ν
⎛ ⎞ ⎛ ⎞+⎜ ⎟ ⎜ ⎟− ⎝ ⎠ ⎝ ⎠
0.250.75 2.4 0.5
1 XG L BI
B L
Rotation, Kθy' (about y axis) ν
⎡ ⎤⎛ ⎞⎢ ⎥⎜ ⎟− ⎝ ⎠⎢ ⎥⎣ ⎦
0.150.75 3
1 YG LI
B
Table note: 1. See Figure 8.4.2.1-1** for definitions of terms
Table 8.4.2.1-2 Stiffness Embedment Factors for a Rigid Plate on a Semi-Infinite Homogeneous
Elastic Half-Space (Adapted from Gazetas, 1991)1
Stiffness Parameter Embedment Factors, ei
Vertical Translation, ez
( )⎡ ⎤⎛ ⎞+⎡ ⎤⎛ ⎞ ⎢ ⎥+ + + ⎜ ⎟⎢ ⎥⎜ ⎟ ⎜ ⎟⎢ ⎥⎝ ⎠⎣ ⎦ ⎝ ⎠⎣ ⎦
0.672 2
1 0.095 1 1.3 1 0.2L BD B d
B L LB
Horizontal Translation, ey (toward long side)
( )⎧ ⎫⎡ ⎤⎛ ⎞− +⎪ ⎪⎜ ⎟⎢ ⎥⎡ ⎤⎛ ⎞ ⎪ ⎪⎝ ⎠⎢ ⎥+ +⎢ ⎥ ⎨ ⎬⎜ ⎟ ⎢ ⎥⎝ ⎠⎢ ⎥ ⎪ ⎪⎣ ⎦ ⎢ ⎥⎪ ⎪⎣ ⎦⎩ ⎭
0.4
0.5
2
162 21 0.15 1 0.52
dD L B dDB B L
Horizontal Translation, ex (toward short side)
( )⎧ ⎫⎡ ⎤⎛ ⎞− +⎪ ⎪⎜ ⎟⎢ ⎥⎡ ⎤⎛ ⎞ ⎪ ⎪⎝ ⎠⎢ ⎥+ +⎢ ⎥ ⎨ ⎬⎜ ⎟ ⎢ ⎥⎝ ⎠⎢ ⎥ ⎪ ⎪⎣ ⎦ ⎢ ⎥⎪ ⎪⎣ ⎦⎩ ⎭
0.4
0.5
2
162 21 0.15 1 0.52
dD L B dDL L B
Rotation, eθx (about x axis)
−⎛ ⎞⎛ ⎞ ⎛ ⎞+ +⎜ ⎟⎜ ⎟ ⎜ ⎟⎜ ⎟⎝ ⎠ ⎝ ⎠⎝ ⎠
0.20 0.5021 2.52 1d d d BB B D L
Rotation, eθy (about y axis)
−⎛ ⎞⎛ ⎞ ⎛ ⎞ ⎛ ⎞+ +⎜ ⎟⎜ ⎟ ⎜ ⎟ ⎜ ⎟⎜ ⎟⎝ ⎠ ⎝ ⎠ ⎝ ⎠⎝ ⎠
0.60 1.9 0.602 21 0.92 1.5d d dL L D
Table note: Embedment factors multiplied by spring
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y
x
y
x
z
z
Plan
Section
L (length)
B (width)
d (thickness)
D (depth)
Homogeneous Soil Properties G (shearing modulus)
ν ( Poisson's ratio)
Figure 8.4.2.1-1 Properties of a Rigid Plate on a Semi-Infinite Homogeneous Elastic Half-Space
for Stiffness Calculations
8.4.2.2 Moment-Rotation and Shear-Displacement Relationships for Footing (Nonliquefiable Sites)
The moment and shear capacity of the foundation shall be confirmed for design loads given in Article 4.8. Moment-rotation and shear force-displacement relationships shall be developed as required by Article 5.3.4. Unless approved otherwise by the owner, the moment-rotation curve for SDAP E shall be represented by a bilinear, moment-rotation curve. The initial slope of the bi-linear curve shall be defined by the rotational spring constant given in Article 8.4.2.1.
The maximum resisting force (i.e., plastic capacity) on the force-deformation curve shall be defined for the best-estimate case. The footing liftoff shall be no more than 50% of the footing area at peak displacement during the push-over analysis, unless special studies are performed and approved by the owner. A bilinear force displacement relationship shall
also be developed for the shear component of resistance.
This approach shall not be used at sites that will liquefy during seismic loading. See Article 8.4.2.3 for sites that liquefy.
8.4.2.3 Liquefaction and Dynamic
Settlement
An evaluation of the potential for liquefaction within near-surface soil shall be made in accordance with requirements given in Article 8.6 and Appendix D of these specifications. If liquefaction is predicted to occur under the design ground motion, spread footings foundations shall not be used unless
• the footing is located below the liquefiable layer,
• ground improvement is performed to mitigate the occurrence of liquefaction, or
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• special studies are conducted to demonstrate that the occurrence of liquefaction will not be detrimental to the performance of the bridge support system.
The owner’s approval shall be obtained before initiating ground improvement or special studies.
8.4.3 Driven Piles
8.4.3.1 General
Resistance factors for pile capacities shall be as specified in Table 10.5.4-2 of the AASHTO LRFD provisions, with the exception that resistance factors of 1.0 shall be used for seismic loads.
For the effect of settling ground and downdrag loads, unfactored load and resistance factors (γ = 1.0; φ = 1.0) shall be used, unless required otherwise by the owner.
Batter piles shall not be used where downdrag loads are expected unless special studies are performed.
For seismic loading the groundwater table location shall be the average groundwater location, unless the owner approves otherwise.
8.4.3.2 Design Requirements
The design of driven pile foundations shall be based column loads determined by capacity design principles (Article 4.8) or elastic seismic forces, whichever is smaller. Both the structural and geotechnical elements of the foundation shall be designed for the capacity design forces of Article 4.8.
Foundation flexibility (Article 5.3.4) shall be incorporated into design for Soil Profile Types
C, D, and E, if the effects of foundation flexibility contribute more than 20% to the displacement of the system. For SDAP E foundations flexibility shall be included in the push-over analysis whenever it is included in the dynamic analysis.
Liquefaction shall be considered when applicable during the development of spring constants and capacity values for these seismic design and analysis procedures.
8.4.3.3 Axial and Rocking Stiffness for Driven Pile/Pile Cap Foundations (Nonliquefiable Sites)
The axial stiffness of the driven pile foundations shall be determined for design cases in which foundation flexibility is included. For many applications, the axial stiffness of a group of piles can be estimated within sufficient accuracy using the following equation: Ksv = Σ 1.25AE/L (8.4.3.2-1)
where
A = cross-sectional area of the pile E = modulus of elasticity of the piles L = length of the piles N = number of piles in group and is
represented by the summation symbol in the above equations.
The rocking spring stiffness values about each horizontal pile cap axis can be computed assuming each axial pile spring acts as a discrete Winkler spring. The rotational spring constant (i.e., moment per unit rotation) is then given by Ksrv = Σ kvn Sn
2 (8.4.3.2-2)
where
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kvn = axial stiffness of the nth pile Sn = distance between the nth pile and
the axis of rotation
The effects of group action on the determination of stiffness shall be considered if the center-to-center spacing of piles for the group in the direction of loading is closer than 3 pile diameters.
8.4.3.4 Lateral Stiffness Parameters for Driven Pile/Pile Cap Foundations (Nonliquefiable Sites)
The lateral stiffness parameters of driven pile foundations shall be estimated for design cases in which foundation flexibility is included. Lateral response of a pile foundation system depends on the stiffness of the piles and, very often, the stiffness of the pile cap. Procedures for defining the stiffness of the pile component of the foundation system are covered in this article. Methods for introducing the pile cap stiffness are addressed in Article 8.4.3.5.
For preliminary analyses involving an estimate of the elastic displacements of the bridge, pile stiffness values can be obtained by using a series of charts prepared by Lam and Martin (1986). These charts are reproduced in Figures 8.4.3.4-1 through 8.4.3.4-6. The charts are applicable for mildly nonlinear response, where the elastic response of the pile dominates the nonlinear soil stiffness.
For push-over analyses the lateral load displacement relationship must be extended into the nonlinear range of response. It is usually necessary to use computer methods to develop the load-displacement relationship in
this range, as both the nonlinearity of the pile and the soil must be considered. Programs such as LPILE (Reese and Wang, 1997), COM 624 (Wang and Reese, 1991), and FLPIER (Hoit and McVay, 1996) are used for this purpose. These programs use nonlinear "p-y" curves to represent the load-displacement response of the soil; they also can accommodate different types of pile-head fixity. Procedures for determining the "p-y" curves are discussed by Lam and
Martin (1986) and more recently by Reese et al. (1997).
The effects of group action on lateral stiffness shall be considered if the center-to-center spacing of the piles is closer than 3 pile diameters.
8.4.3.5 Pile Cap Stiffness and Capacity
The stiffness and capacity of the pile cap shall be considered in the design of the pile foundation. The pile cap provides horizontal resistance to the shear loading in the column. Procedures for evaluating the stiffness and the capacity of the footing in shear shall follow procedures given in Article C8.4.2.2 for spread footings, except that the base shear resistance of the cap shall be neglected.
When considering a system comprised of a pile and pile cap, the stiffness of each shall be considered as two springs in parallel. The composite spring shall be developed by adding the reaction for each spring at equal displacements.
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Figure 8.4.3.4-1 Recommendations for Coefficient of Variation in
Subgrade Modulus with Depth for Sand (ATC, 1996)
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Figure 8.4.3.4-2 Recommendations for Coefficient of Variation in Subgrade Modulus with Depth for Clay (ATC, 1996)
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Figure 8.4.3.4-3 Coefficient of Lateral Pile Head Stiffness for Free-Head Pile Lateral Stiffness
(ATC, 1996)
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Figure 8.4.3.4-4 Coefficient for Lateral Pile-Head Stiffness for Fixed-Head Pile Lateral Stiffness
(ATC, 1996)
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Figure 8.4.3.4-5 Coefficient for Pile Head Rotation (ATC, 1996)
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Figure 8.4.3.4-6 Coefficient for Cross-Coupling Stiffness Term (ATC, 1996)
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8.4.3.6 Moment and Shear Design (Nonliquefiable Sites)
The capacity of the structural elements of driven pile foundations shall be designed to resist the capacity design forces of Article 4.8 or the elastic design force within the column, whichever is smaller. Unfactored resistance (φ = 1.0) shall be used in performing the geotechnical capacity check. The load on the leading row of piles during overturning shall not exceed the plunging capacity of the piles. Separation between the pile tip and the soil (i.e. gapping) shall be allowed only in the most distant row of trailing piles. Forces on all other rows of piles shall either be compressive or not exceed the nominal tension capacity of the piles. The maximum shear force on the pile(s) shall be less than the structural shear capacity of the piles.
If the plunging capacity is exceeded or gapping of other than the trailing row of piles occurs, special studies shall be conducted to show that performance of the pile system is acceptable. Special studies shall be performed only with the prior consent of the owner and require SDAP E.
8.4.3.7 Liquefaction and Dynamic Settlement Evaluations
If liquefaction is predicted to occur at the site, effects of liquefaction on the bridge foundation shall be evaluated. This evaluation shall consider the potential for loss in lateral bearing support, flow and lateral spreading of the soil, settlement below the toe of the pile, and settlement from drag loads on the pile as excess porewater pressures in liquefied soil dissipate. Procedures given in Appendix D shall be followed when making these evaluations.
If liquefaction causes unacceptable bridge performance, consideration should be given to
the use of ground improvement methods to meet design requirements. In light of the potential costs of ground improvement, the owner shall be consulted before proceeding with a design for ground improvement to review the risks associated with liquefaction relative to the costs for remediating the liquefaction potential.
8.4.4 Drilled Shafts
Procedures identified in Article 8.4.3, including those for liquefaction and dynamic settlement, generally apply with the exceptions that, (1) the ultimate capacity of single shaft foundations in compression and uplift shall not be exceeded under maximum seismic loads and (2) the flexibility of the drilled shaft shall be represented using either the estimated depth of fixity or soil springs in a lateral pile analysis.
Checks shall be conducted to confirm that minimum shaft lengths occur. The stable length can be determined by conducting nonlinear computer modeling or by using a length (L) > πλ where λ = [EIp/Es]0.25 for cohesive soils, and
λ = [EIp/f] 0.20 for cohesionless soils
where E = Young’s modulus of the shaft
Ip = moment of inertia of the shaft
F = coefficient of variation of subgrade modulus
Es = subgrade modulus of soil
Z = embedded depth of the shaft
The nonlinear properties of the shaft shall be considered in evaluating the lateral response of the pile to lateral loads during a seismic event. Diameter adjustments shall be
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considered during lateral analyses of shafts with a diameter greater than 600 mm if the shaft is free to rotate, as in the case of a column extension (i.e., no pile cap). Contributions from base shear shall also be considered.
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8.6 LIQUEFACTION DESIGN REQUIREMENTS
8.6.1 General
An evaluation of the potential for and consequences of liquefaction within near-surface soil shall be made in accordance with the following requirements: A liquefaction assessment is required unless one of the following conditions is met or as directed otherwise by the owner. • Mean magnitude for the MCE event is less
than 6.0 (Figures 8.6.1-1 to 8.6.1-4); • Mean magnitude of the MCE event is less than
6.4 and equal to or greater than 6.0, and the normalized Standard Penetration Test (SPT) blow count [(N1)60] is greater than 20;
• Mean magnitude for the MCE event is less than 6.4 and equal to or greater than 6.0, (N1)60 is greater than 15, and FaSs is between 0.25 and 0.375.
If the mean magnitude shown in Figures 8.6.1-1 to 8.6.1-4 is greater than or equal to 6.4, or if the above requirements are not met for magnitudes between 6.0 and 6.4 or if for the Expected Earthquake, FaSs is greater than 0.375, evaluations of liquefaction and associated phenomena such as lateral flow, lateral spreading, and dynamic settlement shall be evaluated in accordance with these Specifications.
8.6.2 Evaluation of Liquefaction Potential
Procedures given in Appendix D shall be used to evaluate the potential for liquefaction.
8.6.3 Evaluation of the Effects of Liquefaction and Lateral Ground Movement
Procedures given in Appendix D shall be used to evaluate the potential for and effects of liquefaction and liquefaction-related permanent ground movement (i.e., lateral spreading, lateral flow, and dynamic settlement). If both liquefaction and ground movement occur, they shall be treated as separate and independent load cases, unless agreed to or directed otherwise by the owner.
8.6.4 Design Requirements if Liquefaction and Ground Movement Occurs
If it is determined from Appendix D that liquefaction can occur at a bridge site, then one or more of the following approaches shall be implemented in the design.
Bridges shall be supported on deep foundations unless (1) the footing is located below the liquefiable layer, (2) special design studies are conducted to demonstrate that the footing will tolerate liquefaction, or (3) the ground is improved so that liquefaction does not occur. If spread footings are being considered for use at a liquefiable site, owner approval shall be obtained before beginning the design process.
If liquefaction occurs, then the bridge shall be designed and analyzed in two configurations as follows: 1. Nonliquefied Configuration: The structure
shall be analyzed and designed, assuming no liquefaction occurs using the ground response spectrum appropriate for the site soil conditions.
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Figure 8.6.1-1 Mean Earthquake Magnitude Map for Western United States
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Figure 8.6.1-2 Mean Earthquake Magnitude Map for Eastern United States
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Figure 8.6.1-3 Mean Earthquake Magnitude Map for Alaska
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Figure 8.6.1-4 Mean Earthquake Magnitude Map for Southeast Alaska
PART I: SPECIFICATIONS 2003 GUIDELINES FOR THE SEISMIC DESIGN OF HIGHWAY BRIDGES
MCEER/ATC-49 124 SECTION 7
2. Liquefied Configuration: The structure as designed in Nonliquefied Configuration above shall be reanalyzed and redesigned, if necessary, assuming that the layer has liquefied and the liquefied soil provides whatever residual resistance is appropriate (i.e., “p-y curves” or modulus of subgrade reaction values for lateral pile response analyses consistent with liquefied soil conditions). The design spectra shall be the same as that used in Nonliquefied Configuration unless a site-specific response spectra has been developed using nonlinear, effective stress methods (e.g., computer program DESRA or equivalent) that properly account for the buildup in pore-water pressure and stiffness degradation in liquefiable layers. The reduced response spectra resulting from the site-specific nonlinear, effective stress analyses shall not be less than 2/3’s of that used in Nonliquefied Configuration. The Designer shall provide a drawing of the load path and energy dissipation mechanisms in this condition as required by Article 3.3 since it is likely that plastic hinges will occur in different locations than for the non-liquefied case. Shear reinforcement given in Article 8.8.2.3 shall be used in all concrete and prestressed concrete piles to a depth of 3 pile diameters below the liquefied layer.
If lateral flow or lateral spreading occurs, the following options shall be considered. 1. Design the piles to resist the forces
generated by the lateral spreading.
2. If the structure cannot be designed to resist the forces, assess whether the structure is able to tolerate the anticipated movements and meet the geometric and structural constraints of Table C3.2-1. The maximum plastic rotation of the piles is 0.05 radians as per Article 8.7.9 and 8.8.6.
3. If the structure cannot meet the performance requirements of Table 3.2-1, assess the costs and benefits of various mitigation measures to minimize the movements to a tolerable level to meet
the desired performance objective. If a higher performance is desired so that the piles will not have to be replaced, the allowable plastic rotations in-ground hinges of Article 8.7.9.2 and 8.8.6.2 shall be met.
8.6.5 Detailed Foundation Design Requirements
Article 8.4 contains detailed design requirements for each of the different foundation types.
8.6.6 Other Collateral Hazards
The potential occurrence of collateral hazards resulting from fault rupture, landsliding, differential ground compaction, and flooding and inundation shall be evaluated. Procedures for making these evaluations are summarized in Appendix D.
1325 NCHRP 20-7(193) Task 6 Report.doc 5B-1
TASK 5
APPENDIX 5B
ASSESSMENT AND MITIGATION OF LIQUEFACTION HAZARDS TO BRIDGE
APPROACH EMBANKMENTS IN OREGON
1325 NCHRP 20-7(193) Task 6 Report.doc
ASSESSMENT AND MITIGATION OF
LIQUEFACTION HAZARDS TO
BRIDGE APPROACH EMBANKMENTS
IN OREGON
Final Report
SPR 361
by
Dr. Stephen E. Dickenson Associate Professor
and Nason J. McCullough
Mark G. Barkau Bryan J. Wavra
Graduate Research Assistants Dept. of Civil Construction and Environmental Engineering
Oregon State University Corvallis, OR 97331
for
Oregon Department of Transportation
Research Group 200 Hawthorne Ave. SE Salem, OR 97301-5192
And
Federal Highway Administration
Washington, D.C. 20590
November 2002