NUMERICAL AND EXPERIMENTAL INVESTIGATIONS OF …

NUMERICAL AND EXPERIMENTAL INVESTIGATIONS OF CONNECTION

FOR TIMBER-STEEL HYBRID SYSTEM

by

Md Riasat Azim

B.Sc., Bangladesh University of Engineering & Technology, 2011

A THESIS SUBMITTED IN PARTIAL FULFILLMENT OF

THE REQUIREMENTS FOR THE DEGREE OF

MASTER OF APPLIED SCIENCE

in

THE FACULTY OF GRADUATE AND POSTDOCTORAL STUDIES

(Civil Engineering)

THE UNIVERSITY OF BRITISH COLUMBIA

(Vancouver)

August 2014

© Md Riasat Azim, 2014

ii

Abstract

In recent years, hybrid systems have grown in popularity as potential solution for mid-rise

construction. There is also an increased interest in using timber for such systems. The lack of

established design guidance, however, has tabled the practical implementation of timber-based

hybrid structures. The aim of this thesis is to address the existing knowledge gap regarding the

detailed connection design of hybrid systems through combined experimental and numerical

investigations on a novel timber-steel system called “FFTT”. The FFTT system relies on wall

panels of mass timber such as Cross-Laminated-Timber (CLT) for gravity and lateral load

resistance and embedded steel beam sections to provide ductility under seismic loading. A vital

step towards practical implementation of the FFTT system is to obtain the proof that the

connections facilitate the desired ‘strong column – weak beam’ failure mechanism.

The numerical work applied the software ANSYS; a parametric study based on the results of

previous tests was conducted to obtain a suitable connection configuration for improved

structural performance. The experimental work, carried out at FPInnovations, consisted of

quasi-static monotonic and reversed cyclic tests on two different connection configurations:

fully and partially embedded ASTM wide flange sections in combination with 7 ply CLT

panels. The combination of partial embedment length and full embedment depth, even when

using the smallest wide flange section, did not facilitate the desired behavior. The connection

performance was significantly improved when reducing the embedment depth (to avoid creating

stress peaks on a weak cross layer) and increasing the embedment length (larger center to center

distance between bearing plates). The used small size steel beam, however, is not practical for a

real structure; therefore, further studies with larger beams and a modified geometry are

recommended before the FFTT system can be applied in practice.

iii

Preface

The numerical analysis section of chapter 3 has been accepted for publication at the proceedings

of World Conference in Timber Engineering:

“Bhat, P., Azim, M.R., Tannert, T., Popovsky, M. “Experimental and numerical investigation of

novel steel-timber-hybrid system”, Proceedings of World Conference in Timber Engineering,

Quebec City, August 10-14, 2014.”

I conducted the numerical studies and wrote that portion of the manuscript. The main section on

“Experimental Investigation” was drafted by Bhat and revised by Tannert and Popovski.

iv

Table of Contents

Abstract .................................................................................................................................... ii

Preface ..................................................................................................................................... iii

Table of Contents.................................................................................................................... iv

List of Figures ......................................................................................................................... ix

Acknowledgements ............................................................................................................... xiii

Dedication.............................................................................................................................. xiv

Chapter 1: INTRODUCTION .............................................................................................. 1

1.1 Tall Timber Structures: Timber-Steel Hybridization ................................................. 1

1.2 Research Need ............................................................................................................ 2

1.3 Research Objective ..................................................................................................... 3

Chapter 2: LITERATURE REVIEW .................................................................................. 4

2.1 Timber and Steel as Structural Materials .................................................................... 4

2.1.1 Cross-Laminated-Timber .................................................................................... 6

2.1.2 Material Modelling of CLT ................................................................................. 8

2.2 Hybrid Construction ................................................................................................. 10

2.2.1 Component Level Hybridization ....................................................................... 10

2.2.2 System Level Hybridization .............................................................................. 11

2.2.3 Hybrid Connections ........................................................................................... 11

2.3 Seismic Force Resisting System Design Principles.................................................. 12

2.3.1 Force-Based Design Approach .......................................................................... 12

2.3.2 Displacement-Based Design Approach ............................................................. 14

2.3.3 Selection of Design Strategy for Hybrid Systems ............................................. 15

v

2.3.4 Capacity Design Concept .................................................................................. 16

2.4 Lateral Load Resisting Systems for Timber Steel Hybrid Structures ....................... 16

2.4.1 Overview of Lateral Load Resisting Systems ................................................... 16

2.4.2 Infill Wall Systems ............................................................................................ 18

2.4.3 Mass Timber Construction ................................................................................ 19

2.5 Recent Experimental Research on CLT and Hybrid Systems .................................. 19

2.5.1 Ceccotti et al (2010) .......................................................................................... 19

2.5.2 Popovski & Karacabeyli (2011) ........................................................................ 20

2.5.3 Fragiacomo et al (2011) .................................................................................... 21

2.5.4 Numerical Investigations on Hybrid Systems ................................................... 22

2.6 FFTT System ............................................................................................................ 24

2.6.1 Structural System .............................................................................................. 24

2.6.2 Experimental Investigations on FFTT Connection ........................................... 26

Chapter 3: NUMERICAL INVESTIGATION ON FFTT SYSTEM .............................. 29

3.1 Finite Element Model Development ......................................................................... 29

3.1.1 Modelling of CLT Panels .................................................................................. 29

3.1.2 Modelling of Steel Beams ................................................................................. 30

3.1.3 Contact Simulation ............................................................................................ 30

3.1.4 Boundary Conditions ......................................................................................... 30

3.1.5 Load Application ............................................................................................... 31

3.1.6 Post Processing .................................................................................................. 31

3.2 Numerical Results ..................................................................................................... 32

3.2.1 Configuration 1: Partially Embedded Wide Flange Section ............................. 32

3.2.2 Configuration 2: Fully Embedded Wide Flange Section .................................. 35

3.2.3 Configuration 3: Fully Embedded Section with Reduced Cross Section .......... 38

vi

3.2.4 Configuration 4: Full Embedment Length of Hollow Steel Section ................. 40

3.2.5 Configuration 5: Reduced Embedment Length of Hollow Steel Section ......... 42

3.2.6 Summary on Model Results from Previous Tests ............................................. 45

3.3 Numerical Study to Improve Connection Configuration .......................................... 45

3.3.1 Geometry ........................................................................................................... 46

3.3.2 Parameter Variation ........................................................................................... 48

3.3.3 Parametric Study Results ................................................................................... 48

3.3.4 Parametric Study with Partial Embedment Depth ............................................. 54

3.4 Discussion on Numerical Analysis and Optimization Studies ................................. 56

Chapter 4: EXPERIMENTAL INVESTIGATION ON FFTT SYSTEM ...................... 63

4.1 Introduction............................................................................................................... 63

4.2 Materials ................................................................................................................... 63

4.3 Specimen Description ............................................................................................... 64

4.4 Test Procedure .......................................................................................................... 66

4.5 Experimental Results ................................................................................................ 68

4.5.1 Series 1: Monotonic Test on Fully Embedded Beam ........................................ 68

4.5.2 Series 1: Cyclic Test on Fully Embedded Beam ............................................... 72

4.5.3 Series 2: Monotonic Test on Partially Embedded Beam ................................... 76

4.5.4 Series 2: Cyclic Test on Fully Embedded Beam ............................................... 78

4.6 Discussion on Experimental Investigations .............................................................. 83

4.6.1 Comparison between Experimental and Numerical Results ............................. 83

4.6.2 Point of Rotation of Beam ................................................................................. 84

4.6.3 Ductility and Force Modification Factor ........................................................... 85

4.6.4 Hysteretic Behavior ........................................................................................... 87

4.6.5 Energy Dissipation ............................................................................................ 92

vii

Chapter 5: CONCLUSIONS ............................................................................................... 93

5.1 Summary ................................................................................................................... 93

5.2 Recommendation for Further Studies ....................................................................... 95

References .............................................................................................................................. 96

viii

List of Tables

Table 1: Material Properties of Steel and Structural Timber (Yalda, 2009) .................................. 4

Table 2: Physical properties of CLT (Gagnon & Pirvu, 2011) ...................................................... 8

Table 3: Elastic properties of CLT (Gsell et al., 2007) .................................................................. 9

Table 4: FFTT System Options .................................................................................................... 26

Table 5: Properties of CLT ........................................................................................................... 29

Table 6: Properties of Steel beam ................................................................................................. 30

Table 7: Results from previous experimental tests and numerical simulation ............................. 45

Table 8: Parameter range for numerical study ............................................................................. 48

Table 9: Results of parametric study (Beam: W 150 x 29.8) ....................................................... 49

Table 10: Results of parametric study (Beam: W 130 x 23.8) ..................................................... 50

Table 11: Results of parametric study (Beam: W 100 x 19.3) ..................................................... 51

Table 12: Test specimen description ............................................................................................ 64

Table 13: Comparison between Experimental results and their numerical simulation ................ 83

Table 14: Ductility ratio and force modification factor ............................................................... 86

Table 15: Cyclic tests results ........................................................................................................ 87

ix

List of Figures

Figure 1: Stress-Strain Relationship- Structural Steel .................................................................... 5

Figure 2: The 3 directions for timber properties (Holtz, 2002) ...................................................... 5

Figure 3: Cross-Laminated-Timber ................................................................................................ 7

Figure 4: Nonlinear material model of timber in compression (Grosse and Rautenstrauch, 2004): a) parallel to grain and b) perpendicular to grain ........................................................................... 9

Figure 5: Nonlinear material model of timber in shear and tension (Multiplas, 2013) ................ 10

Figure 6: Component level hybridization (left: filch Beam, right: Glulam with steel plate) ....... 11

Figure 7: Concept of Hybrid Connection (Yalda, 2009) .............................................................. 12

Figure 8: Force–deformation relationship of a typical plastic hinge (ASCE 41, 2006) ............... 15

Figure 9: Hysteretic model at near-collapse (Ceccotti & Karacabeyli, 2002) ............................. 17

Figure 10: Masonry infill walls Model (Yousuf & Bagchi, 2009) ............................................... 18

Figure 11: CLT Wall Response to Lateral Loading (Schneider, 2009) ....................................... 20

Figure 12: Semi-static CLT Wall Tests - Effect of Connection between Panels: (left) single panel CLT wall, (right) three panel CLT wall (Popovski and Karacabeyli 2011) ....................... 21

Figure 13: 7 Story CLT Shake Table Test (Fragiacomo et al. 2011) ........................................... 22

Figure 14: Solid Panel Core and Intersecting Ductile Steel Link Beams (Green and Karsh, 2012) ...................................................................................................................................................... 24

Figure 15: Type 3 Lateral Load Resisting System for FFTT (Green and Karsh, 2012) .............. 25

Figure 16: Type 3 Lateral Load Resisting System for FFTT (Green and Karsh, 2012) .............. 25

Figure 17: Typical Setup and Instrumentation (Bhat, 2013) ........................................................ 27

Figure 18: Load-deformation plot of the test configuration 4 (Bhat, 2013) ................................. 28

Figure 19: Finite Element Model of test configuration 1 ............................................................. 31

Figure 20: Shear stress plot test configuration 1 .......................................................................... 33

Figure 21: Compressive stress plot test configuration 1 .............................................................. 33

Figure 22: Comparative load deformation plot of test configuration 1: embedded portion ......... 34

x

Figure 23: Comparative load deformation plot of test configuration 1: cantilever portion ......... 34

Figure 24: Finite element model of test configuration 2 .............................................................. 35





Figure 29: Test configuration 3 .................................................................................................... 38









Figure 38: Comparative load deformation plot of test configuration 5 ........................................ 44

Figure 39: Finite element model for numerical optimization 1.................................................... 47

Figure 40: Details of the steel beam with bearing and side plates ............................................... 47

Figure 41: Load-deformation plot at different points of interest for the optimization study when the W100 x19.3 beam was fully embedded with 150 mm bearing length and 350 mm spacing . 53

Figure 42: Contour plot of compressive stress parallel to grain inside the CLT panel when the W100x19.3 beam was fully embedded with 150 mm bearing length and 350 mm spacing ........ 53

Figure 43: Finite element model for numerical optimization 2.................................................... 54

Figure 44: Contour plot of compressive stress parallel to grain numerical model when the W100 x19.3 beam was partially embedded with 150 mm bearing length and 665 mm spacing ............ 55

Figure 45: Load-deformation plot at different points of interest for the optimization study when the beam was W100 x19.3 with 150 mm bearing length and 665 mm spacing ........................... 56

xi

Figure 46: Variation in compressive stress with embedment length of beam (Beam: W 100 x 19.3) .............................................................................................................................................. 60

Figure 47: Variation in compressive stress with length of bearing plate (Beam: W 100 x 19.3) 60

Figure 48: Variation in shear stress with embedment length of beam (Beam: W 100 x 19.3) .... 61

Figure 49: Variation in shear stress with length of bearing plate (Beam: W 100 x 19.3) ............ 61

Figure 50: Variation in displacement with embedment length of beam (Beam: W 100 x 19.3) . 62

Figure 51: Variation in displacement with length of bearing plate (Beam: W 100 x 19.3) ......... 62

Figure 52: Experimental setup for test series 1 ............................................................................ 65

Figure 53: Full embedment of the steel beam inside the CLT panel for test series 1 .................. 65

Figure 54: Full embedment of the steel beam inside the CLT panel for test series 2 .................. 66

Figure 55: CUREE loading protocol for series 1 ......................................................................... 67

Figure 56: CUREE loading protocol for series 2 ......................................................................... 68

Figure 57: Yielding of beam during experimental series 1 .......................................................... 69

Figure 58: Deformation inside the CLT panel during experimental series 1 ............................... 70

Figure 59: Load-displacement curve: Series-1, monotonic test-1 ................................................ 71


Figure 61: Rolling shear failure in CLT panel during cyclic test of series 1 ............................... 73

Figure 62: Cyclic test: Series-1, LVDT-1 .................................................................................... 74



Figure 65: Yielding of beam during experimental series 1 .......................................................... 76



Figure 68: Out of plane buckling of the steel beam during cyclic test of series 2 ....................... 79

Figure 69: Damage in the CLT panel during cyclic test of series 2 ............................................. 80

xii





Figure 74: Points of rotation of beams for series 1 and 2 ............................................................. 85

Figure 75: Cyclic test: Series-1, LVDT-1 (with backbone curve) ............................................... 88







Figure 82: Energy dissipation during reverse cyclic tests ............................................................ 92

xiii

Acknowledgements

First of all, I would like to express my sincere gratitude to Dr. Thomas Tannert, my thesis

supervisor, for his valuable guidance, support and encouragement throughout my graduate

studies. It has been a pleasure and honor to work under his supervision in this project.

I would like to thank Dr. Marjan Popovski from FP Innovations for his constant support and

valuable suggestions. I also acknowledge Mr. Paul Simons, who through his time and effort

made the experimental investigation a success.

I extend my gratitude to Johannes Schneider, whose fabrication skills and support was very

valuable for this research project. Also, the UBC technicians Mark Rigolo, George Lee and

Harald Schrempp were most helpful at different stages of my work.

I thank my fellow MASc students Michael Fairhurst and Alexandra Cheng from the Department

of Civil Engineering for helping me during experimental investigation and proof reading my

thesis, respectively.

I would like to thank Structurlam for providing the timber products for the experimental

program.

Finally, I acknowledge NSERC for the financial support provided through the NewBuildS

network.

xiv

Dedication

To my loving parents and my brother, without their support, I could not

have achieved anything.

1

Chapter 1: INTRODUCTION

1.1 Tall Timber Structures: Timber-Steel Hybridization

Hybrid construction combines the structural and architectural features of components made from

different materials. In hybrid construction, various materials may work independently or act

together, in such a way that they combination is advantageous compared to either single material.

During the last decade, much research has been conducted on applications of hybrid structures;

the information on and details for steel and wood hybrid structures, however, are dispersed and

not readily accessible to builders. As part of this thesis, a literature study on existing hybrid steel

and wood structural systems was conducted to identify current techniques of hybridization along

with the benefits and challenges associated with them. The literature review has highlighted the

opportunity for wood-steel hybrid buildings and existing knowledge gaps.

Tall wood buildings are not a new concept: 19 story wooden pagodas were built in Japan 1400

years ago and are still standing in one of the highly seismic regions in the world. The Stadthaus

project, London (2008) is an example of an innovative system; it is a nine story building

constructed entirely with timber. Its structural system is made of a Cross-Laminated Timber

(CLT), which offers an effective solution for construction of large-scale and tall wood buildings.

In North America, however, the use of structural wood in construction of new high-rise

buildings is not common. History of losses due to fire has regulated the limitations on the

building area and height for timber structures in various structural building codes. In recognition

of improved fire-fighting measures, the BC Building Code (BCBC, 2009) allows the

construction of light-frame wood structures to a maximum of six storeys since 2009; before that

the limit was four storeys.

2

A highly ductile material such as steel, when combined with timber, can enhance the post-yield

behavior of timber structures. A good engineering design of a hybrid system that combines the

merits of the two materials can overcome the limitations of light-frame wood construction and

revoke the building height restrictions currently placed on timber buildings. During the past few

years, extensive research has targeted the construction of timber-based hybrid structures in order

to increase their performance and also, owing to the demands of sustainable construction. One

such system is the FFTT system (Green and Karsh, 2012), which is predominantly a mass-

timber vertical system with embedded steel beam sections that provide ductility in the system.

This system is discussed in detail in Chapter 2.

1.2 Research Need

Mass timber and steel hybrid systems have the potential to impact the building industry, address

issues of climate change and pose a challenge to concrete and steel structures. However, the

current building codes provide no guidelines on seismic design and parameters for the

construction of hybrid systems. Due to lack of design values and guidelines and understanding of

the global behaviour of hybrid systems, the implementation of a large scale timber-steel hybrid

system has not yet been possible in Canada. Analytical and experimental studies that verify the

system performance, identify the challenges and optimize the connections for hybrid systems are

necessary in order to establish design guidelines and enable implementation.

Recently, through collaboration between the University of British Columbia Vancouver (UBC)

and FPInnovations, experimental investigations have been carried at the component level of the

FFTT system. Different connection configurations were tested using quasi-static monotonic as

well as reversed cyclic loading. Though these tests provided valuable information, they need to

3

be complemented by numerical analyses. Numerical modelling of timber can be very complex

due to the fact that the material is anisotropic and that its behavior varies with the type of loading

and also with the direction of loading. Design codes use only elastic properties of timber for

structural design purposes; however, as will be discussed later, timber does exhibit good post

yield inelastic behavior in compression and, consequently, nonlinear modelling can better capture

the system behavior when the structure is subjected to overload (wind and earthquake), which an

elastic model cannot accurately predict. Only considering the elastic properties of timber is

conservative for design. Therefore, there is a need for conducting non-linear numerical

investigations on the FFTT system to understand its behavior at the component level.

1.3 Research Objective

The purpose of this study is to investigate numerically and experimentally the component level

behaviour of the FFTT system and propose a connection layout that can facilitate its successful

implementation in mid-rise and high-rise wood-hybrid structures.

The numerical investigations, described in Chapter 3, complement the results from previous

experiments and improve the connection layout for further experiments. These experiments, as

described in Chapter 4, include monotonic and cyclic loading tests. Based on the numerical and

experimental results, conclusions are drawn and future research needs are outlined in Chapter 5.

4

Chapter 2: LITERATURE REVIEW

2.1 Timber and Steel as Structural Materials

The response of a structural system is influenced by the nature and behavior of the used

construction materials. Hence, for designing timber-steel hybrid structures, it is important to

understand the properties of the individual materials and their potential incompatibility. The

properties of timber vary considerably with species. The properties of Spruce Pine SS (as

representative of timber) and steel, used as construction materials, are summarized in Table 1.

Table 1: Material Properties of Steel and Structural Timber (Yalda, 2009)

Material Density (kg/m3)

Elastic Modulus (MPa)

Compressive Strength (MPa)

Tensile Strength (MPa)

Steel 7,800 200,000 400-1000 400-1000

Spruce Pine SS

400-500 10,500 Parallel 10

Perpendicular 3

Parallel 6

Perpendicular 1

Steel is a homogeneous and isotropic material. It has high tension and compression strengths

along with high stiffness and ability to sustain large inelastic deformation without fracture. Steel

exhibits linear stress-strain relationship up to yielding (Figure 1) and a very good post-yield

behavior providing ductility to the system. This linear region is elastic and the slope of the curve

is the elastic modulus of the material. Beyond yielding, stress increases with increasing

deformation due to strain hardening till ultimate strength after which the material fractures.

The in-elastic force-deformation response of a structure depends on the hysteresis response under

cyclic deformation of the structural materials and components due to inelastic behavior. The area

under the hysteresis loop represents the dissipation of energy. Structural steel dissipates great

5

amounts of energy under cyclic loads. If designed efficiently, steel structures exhibit extreme

ductile behavior during an earthquake event.

Timber is an anisotropic material; that is, the mechanical properties vary in three mutually

perpendicular directions: Longitudinal, Tangential and Radial (Figure 2). The strength properties

are strong parallel to grain and weaker across the grain. Timber exhibits ductile failure in

compression and brittle failure in tension and shear.

Figure 1: Stress-Strain Relationship- Structural Steel

Figure 2: The 3 directions for timber properties (Holtz, 2002)

6

There are inherent uncertainties in the structural properties of timber. Wood is a hygroscopic

material, loss and gain of moisture affects its dimensional stability and strength. In addition, the

properties are dependent on the species and characteristics of the tree from which the timber was

harvested. The growing conditions and local imperfections (like knots) have an impact on the

strength properties (Keenan, 1986). Therefore, engineers use conservative strength properties

based on timber grades as specified in CSA 086 (CSA, 2010). The stress-strain relationship of

wood under compression is non-linear with good post yield behavior under compressive loading.

When subjected to tensile or shear forces, however, timber exhibits brittle failure. Unlike steel,

no cyclic energy dissipation can be observed for structural wood when loaded in tension or shear.

2.1.1 Cross-Laminated-Timber

CLT is a relatively new product which is gaining in popularity in Europe and recently also in

North America. CLT panels are usually made of an odd number of wood layers glued together in

a cross-layer pattern, where each layer is oriented in alternating 90 degree angles. CLT panels

are generally made of three, five, seven etc. layers of softwood glued together (Figure 3). The

gluing is done along the full surface of each panel. Panels are usually manufactured with their

outer layers oriented in the direction that the CLT is going to span (Gagnon & Pirvu, 2011).

Material properties of CLT, e.g. strength in bending and shear, vary according to manufacturer

and raw materials.

7

Figure 3: Cross-Laminated-Timber

Moisture content generally has a significant effect on wood performance due to shrinkage.

Moisture content at delivery for CLT is typically 8–14%. Surface quality of CLT is important for

architectural features and structural use. The estimation of design properties of the CLT not only

depends on the species and quality of wood used, but also the number, orientation, and thickness

of the layers. Classification of the surface quality of the panels is as in following:

• Non-visible Grade: The surface is planned. Such panels are suitable for lining.

• Residential visible: The surface is planned and sanded. These panels are suitable for

residential internal exposure.

• Industrial visible: The surface is planned and lightly sanded. Such panels are suitable for

exposed industrial internal structure.

CLT is generally manufactured with the properties as shown in Table 2.

8

Table 2: Physical properties of CLT (Gagnon & Pirvu, 2011)

Width up to 4 m

Length up to 16 m

Thickness 19 mm, 27.5 mm, 35 mm and 42 mm

Pre-cutting Any cuts for windows, doors and so on

Wood types Spruce (Pine and Larch on request)

Grading C24/C16 (in line with DIN 4074); higher grades on request

Moisture content 12% +/- 2%

Adhesive Formaldehyde free adhesive for edge and surface bonding, finger jointing

Optical qualities Standard and visible quality

Surface finish Sanded

2.1.2 Material Modelling of CLT

The modeling of material properties for CLT is complex, owing to the fact that these properties

vary with species and quality of wood, number of individual layers, their orientation and

thicknesses. For the purpose of numerical modelling, CLT is often considered as a linear elastic

orthotropic material. The various properties of the panel are obtained from experimentation or by

using engineering theorems like Gamma Method, Shear Analogy or Composite Theory.

According to Gsell et al. (2007), the assumption of linear elastic orthotropic material behavior of

CLT is accurate enough to evaluate strength and stiffness properties of panels. The CLT

properties as derived from their study are shown in Table 3 where the x, y, and z subscripts refer

to three mutually orthogonal directions and E0 and E90 refer to elastic modulus of stiffness

parallel and perpendicular to grain direction. These properties can be used for numerical analyses

if linear elastic orthotropic behavior is considered.

9

Table 3: Elastic properties of CLT (Gsell et al., 2007)

Properties Value (MPa) Properties Value

Ey (E0) 8210 γyx 0.090

Ex (E90) 4630 γzx 0.040

Ez 500 γyz 0.364

Gxz 949 γxy 0.051

Gxy 747 γxz 0.380

Gyz 54 γzy 0.022

Not many studies have been carried out regarding the nonlinear modelling of CLT (and timber in

general) owing to the complex behavior of timber post yielding. Grosse and Rautenstrauch

(2004) proposed a five stage nonlinear material model for timber incorporating degradation as

shown in Figure 4 (a) and (b) for compression parallel and perpendicular to grain, respectively.

The shear and tension behavior is usually modelled as linear elastic as shown in Figure 5

(Multiplas, 2013). Grosse’s procedure can be used to model the post-yield inelastic stress-strain

behavior of CLT. However, as of now, no experimental data is available for CLT to numerically

model such behavior.

Figure 4: Nonlinear material model of timber in compression (Grosse and Rautenstrauch, 2004):

a) parallel to grain and b) perpendicular to grain

10

T

Figure 5: Nonlinear material model of timber in shear and tension (Multiplas, 2013)

2.2 Hybrid Construction

All timber structures, to some extent, are hybrid structures since connections are made using

steel and foundations are usually concrete. However, true hybridization is the process of

combining two or more materials to form a system by making use of the strength of each

material and overcome their weaknesses. Hybridization can be classified as component level and

system level hybridization (Yalda 2009).

2.2.1 Component Level Hybridization

Component level hybridization exists when two different materials are combined together to act

as a single structural unit (Figure 6). Common examples for this hybridization are hybrid bridge

decks, hybrid slab/diaphragms, hybrid columns and hybrid beams (such as flitch beams).

11

Figure 6: Component level hybridization (left: filch Beam, right: Glulam with steel plate)

2.2.2 System Level Hybridization

System hybridization combines different materials at the structural level to share the loads acting

on them. Common examples for this type of hybridization are mixed vertical systems where the

first few stories are built from a material different from that of the upper stories, hybrid roof

trusses where timber is placed at the top of the truss and steel as bottom chord, and hybrid frames

where wood and steel share both gravity and lateral loads. Limited research results are currently

available on the response and behavior of steel-timber hybrid structures.

2.2.3 Hybrid Connections

Due to material and structural differences between steel and wood, and efficient connection

between the materials is of high priority (Figure 7). While combining steel with wood,

dimensional changes like thermal expansion/contraction of steel and wood shrinkage/swelling

may occur with time. Steel plates are commonly used for connections in wood/steel hybrid

structures. Johansen’s yield model (Johansen, 1949) is adopted in CSA 086 (CSA, 2010) for the

design of dowel-type connections. An ideal connection between steel and timber should lead to

yielding of the steel connectors before the wood crushes. Splitting of wood is a brittle failure,

and hence, should be avoided.

12

Figure 7: Concept of Hybrid Connection (Yalda, 2009)

2.3 Seismic Force Resisting System Design Principles

The two main principles of designing Seismic Force Resisting Systems (SFRSs) are Force-Based

Design and Displacement-Based Design.

2.3.1 Force-Based Design Approach

In the force-based design approach, the maximum force experienced by the system is evaluated,

which is the structure’s base shear. This force is reduced by seismic reduction factors accounting

for ductility and over-strength and redistributed proportionally along the height of the building.

Maximum Base Overturning Moments are calculated. The system is then designed to resist these

forces and moments. NBCC 2010 (NRC, 2010) uses the Equivalent Static Force procedure to

determine Base Shear and distribution of story shear. Base shear (Vbase) is calculated using:

,---------------------------------------------------------------(1)

Where Sa(T) is the building acceleration, Mv are higher mode effects, Ie is the importance factor,

Rd is the ductility factor, Ro is the over-strength factor, and W is the weight of the building.

13

The building fundamental period T is estimated using empirical formulae and limits the building

period calculated from analytical model to certain value to account for non-structural

components adding stiffness, model inaccuracies and to ensure minimum strength.

A hybrid of timber-steel is lighter than a regular steel frame structure. With this reduction in

weight, seismic performance of the structure can be enhanced. The elastic forces evaluated are

modified to “Design Forces” by reduction factors namely ductility and over-strength factors.

This approach allows for inelastic deformation in the structure dissipating energy during a

seismic event. NBCC classifies ductility levels into four categories- Ductile (D), Moderately

Ductile (MD), Limited Ductility (LD), and Conventional Construction (CC). Systems with high

ductility have specific design requirements and demand rigorous detailing.

The ductility factor is given by the ratio of ultimate roof drift to yield roof drift:

--------------------------------------------------------------------------------(2)

Where δ is defined as the point of first yield anywhere in SFRS and δ is the ultimate drift (the

deformation at the point of “near collapse”).

The over-strength factor accounts for the available over strength in the system. It is defined as

the ratio of maximum base shear resistance (Vmax) to the design base shear (V) (FEMA, 2009).

--------------------------------------------------------------------------------------(3)

There is no guideline provided so far in NBCC 2010 about the values of Rd and Ro for CLT shear

wall buildings, however as per FPInnovations (Gagnon and Pirvu, 2011), these values can be

conservatively assumed as 2.0 and 1.5, respectively.

14

2.3.2 Displacement-Based Design Approach

The displacement-based design approach evaluates the maximum deformation experienced by

the structure, and the system is the designed to resist this deformation either elastically or

plastically. Plastic design ensures dissipation of energy during a seismic event; it results in larger

deformation in which case the acceptance criteria are set to determine allowable damage in the

structure without leading to collapse. This method is known as Performance Based Plastic

Design (Wang et al., 2011). Performance is defined as the acceptable level of damage in the

system. The estimation of the structural performance involves several uncertainties like variation

in ground motion characteristics and the capacity of the components of the system to resist the

imposed demands. Therefore, performance-based design follows a probabilistic design

philosophy with the probability of exceedance of a certain desired performance.

The performance-based design approach is supported in ASCE 41 (2006) for seismic revaluation

and rehabilitation of structures. Hinges are defined as the point of plastic yielding. Each point on

the hinge behavior model (Figure 8) corresponds to different performance levels that define

acceptance criteria of plastic deformation for each level. Immediate Occupancy (IO) Level

occurs just after plastic yielding (Point B) while Life Safety (LS) level occurs significantly

before point of total collapse (Point C). Prevention of Collapse (CP) Level corresponds to

deformation just before the failure point. For the structure to be operational, the deformation is

expected to be below Point B. Typically, for hinges under bending, the acceptance criterion is

indicated in terms of rotations or curvatures.

15

Figure 8: Force–deformation relationship of a typical plastic hinge (ASCE 41, 2006)

2.3.3 Selection of Design Strategy for Hybrid Systems

Current force-based design procedures use spectral acceleration to determine the lateral strength

required by the system to remain elastic and then applies seismic reduction factors that account

for inherent ductility and over-strength (ASCE 41, 2006). One shortcomings of this approach lies

in the determination of fundamental period of the system. Empirical formulae for elastic

fundamental period available in the design code are not particularly tailored for hybrid systems.

In order to minimize damage in wood frame buildings, inter-story drift can be used as key

parameter for seismic design. Although the limitations of force-based procedure are alleviated,

this approach is not extensively used in the design of timber buildings. This approach requires

knowledge of global nonlinear monotonic load-displacement behavior of the building and

viscous damping at a target displacement. In addition to sophisticated structural analysis models,

system testing is necessary to obtain the required information for the design. With further

research and test results on the global behavior of timber structures, this design procedure can be

proven valuable in controlling damage in timber buildings resulting from seismic events.

16

2.3.4 Capacity Design Concept

Capacity design is a principle that is based on the hypothetical behavior of the structure under

seismic load. The system is designed so as to trigger a desired mechanism during a seismic event

and suppress the undesired response. This behaviour is achieved by predetermining the weak link

in the system and then designing to initiate dissipation of energy by yielding of those members of

higher ductile nature and limit inelastic behavior of other components to avoid potential brittle

failure (Mitchell, et al., 2003). The system is detailed to accommodate large deformations during

an expected duration of strong ground motion without significant loss of lateral strength and

ensuring the integrity of the system to sustain gravity loads.

The main difference between force based design and capacity design is that, in the former a

particular force is calculated and the structure is proportioned to resist that load while for the

latter, the required performance of the structure is known and the force is calculated. This design

philosophy appears to be useful in the design of hybrid structures in order to avoid complex

techniques of determining potential collapse mechanism. This design strategy helps to develop a

hierarchy of capacity among the components of the structure.

2.4 Lateral Load Resisting Systems for Timber Steel Hybrid Structures

2.4.1 Overview of Lateral Load Resisting Systems

In steel structures, moment frames mostly form the primary SFRS, often combined with bracing.

In timber structures, lateral loads are transferred to the foundation by vertical bracing achieved

mainly by shear walls with panel sheathing. Wood moment frames are not usually preferred

since it is difficult to achieve a moment connection between wood members. Hybrid systems

17

could be used to resist the combination of lateral and gravity loads to enhance seismic

performance of the timber structures.

The use of structural panels is one of the most efficient ways of providing lateral support (Dickof

et al., 2012). Plywood and OSB panels can be used for horizontal diaphragms and shear walls to

brace the building for wind and seismic loads. Floor diaphragms are assumed to behave as deep

I-Beams and the supporting shear wall transfers the loads to the foundation. The connections

between the shear walls and diaphragm must be efficiently engineered and the wall should be

anchored adequately to ensure systematic load transfer and avoid overturn under lateral loads.

The performance of timber structures during a seismic event is highly dependent on the behavior

of its connections under cyclic loading. Wood in tension behaves linearly and elastically under

cyclic loads and failure is brittle in nature with no dissipation of energy. Steel connections in

timber structures are designed to be “semi-rigid” connections instead of perfectly rigid allowing

for plastic deformation and energy dissipation. The pinching hysteric model of wood wall system

developed at the University of Florence (Ceccotti & Karacabeyli, 2002), is shown in Figure 9.

The force-deformation curve is initially steep till its elastic limit, and then the curve becomes

non-linear and less steep reaching a peak, where the maximum connection capacity may be

found Fmax. Ultimate displacement at “near collapse” criterion was taken as 0.8Fmax.

Figure 9: Hysteretic model at near-collapse (Ceccotti & Karacabeyli, 2002)

18

2.4.2 Infill Wall Systems

Common infill wall systems include masonry infill walls in steel or concrete moment frames, see

Figure 10. Previous studies have confirmed the increase in stiffness and strength of the frame;

but on the other hand, they also decrease the system ductility (Kodor et al 1995).

Figure 10: Masonry infill walls Model (Yousuf & Bagchi, 2009)

Typically, infill walls are not accounted for in the structural design of the system, but only the

contributing addition mass is considered. However, addition of relatively stiff masonry infill wall

in Ductile Steel Moment Frame has a significant impact on the seismic performance of the

system due to high flexibility of steel frame and high stiffness of masonry walls. Yousuf and

Bagchi (2009) confirmed that infill walls reduce the deflection and ductility in the system and

hinging occurred in columns at locations other than the base. Therefore, it is necessary to isolate

the infill walls from moment frame and be designed as structural components.

Masonry infill walls are typically designed as diagonal struts as shown in Figure 10. The CLT

infill panels are found to provide higher strength and stiffness than OSB/Plywood shear walls.

The reduction in ductility is least severe for low ductility moment frames and no evident benefit

was found in choosing high ductility over low ductility moment frame. More detailed parametric

studies are required to optimize the member sizes in order to get maximum ductility in the

19

system. Further research and experimental testing, mainly seismic reduction factors and

connection behavior, need to be carried out for successful implementation of such a hybrid

system (Dickof et al, 2012).

2.4.3 Mass Timber Construction

Tall wood buildings are not a new concept. One example are Pagodas as high as 19 story

buildings in Japan, built 1400 years ago and still standing in one of the highly seismic regions in

the world (Green and Karsh, 2012). The Stadthaus project, London (2008) is an example of using

mass timber in multi-story construction. It is a nine story building constructed entirely with

timber (CLT), claimed as the world’s tallest pure timber residential building (at the time of

completion). Mass timber construction is an approach of combining mass timber panels with

structural technology to produce a system whose behavior is significantly different from light

wood system. They behave more like concrete structures. Mass timber such as Laminated Strand

Lumber (LSL), Laminated Veneer Lumber (LVL) and CLT are not only stronger and stiffer than

conventional timber but also easier to design with due to their higher uniformity.

2.5 Recent Experimental Research on CLT and Hybrid Systems

2.5.1 Ceccotti et al (2010)

Large scale dynamic tests have been performed to evaluate the ductility and overstrength factor

for CLT panel structures; e.g. a three storey CLT building test was performed on a unidirectional

shake table by NIED and CNR-IVALSA in Japan (Ceccotti et al, 2010). Tests were performed

using Kobe, El Centro, and Nocera Umbra ground motions adjusted to peak ground accelerations

for 0.15 g and 0.5 g. The test building was approximately 7 × 7 m in plan and 10 m tall. The

walls were composed on 85 mm thick wall panels and 142 mm thick floor panels. No damage

20

was observed in any component at a peak ground acceleration of 0.5 g. When the ground

acceleration was increased to 0.8 g, slight deformation was noticed in the screws at the vertical

joints between the panels. Hold down failure was observed through pull out and bending of the

nails when the peak ground acceleration was increased to 1.2 g deformation in the screws

between the panels was also observed.

2.5.2 Popovski & Karacabeyli (2011)

To determine the structural properties of CLT, Popovski and Karacabeyli (2011) performed a

number of semi-static tests on CLT walls. The set up included varying connectors at the base.

Single CLT wall and three panel CLT walls were tested. For both the walls, the height and length

were 2.3 m and 3.45 m, respectively. The three panel wall was step jointed by screwing between

panels. At the base of both walls, Type B brackets of 3.9 mm diameter and 89 mm length were

used. Upon cyclic loading in accordance with CUREE protocol, three types of responses were

observed, overturning, rocking or combination of the two, see Figure 11. Rocking and deflection

of connection caused the maximum energy dissipation and subsequent failure. Hysteretic

pinching behavior was observed as shown in Figure 12.

Figure 11: CLT Wall Response to Lateral Loading (Schneider, 2009)

21

Figure 12: Semi-static CLT Wall Tests - Effect of Connection between Panels: (left) single panel

CLT wall, (right) three panel CLT wall (Popovski and Karacabeyli 2011)

CLT walls showed enhanced seismic performance compared to light frame construction: CLT

construction is far less susceptible to “Soft Story” mechanism than the platform frame systems

since the panels are also vertical loading carrying components and remain in place without

complete collapse (Popovski & Karacabeyli, 2011).

2.5.3 Fragiacomo et al (2011)

A seven story building made of CLT slabs and walls was tested at the E-defense shake table as

shown in Figure 13. The building was subjected to 100% of the Kobe earthquake with a peak

ground acceleration of 0.82 g in one direction and 0.6 g in the perpendicular direction. The

building responded with limited structural damage. Some damage to the connectors in the hold

downs were noticed, although no failure occurred. Additionally, with appropriately ductile

connections between the wall panels, an Rd of 3.0 is achieved (Fragiacomo et al., 2011). This

finding was supported by other research with appropriate ductile connections (Yeoh et al, 2011).

22

Figure 13: 7 Story CLT Shake Table Test (Fragiacomo et al. 2011)

2.5.4 Numerical Investigations on Hybrid Systems

Rinaldin et al. (2011) created a numerical model of a single CLT panel with connecting brackets.

Two non-linear hysteretic behavior models were created for the brackets: one for shear only, to

represent the angle bracket connections, and one for tension and compression in the hold-down

connections. The timber panel was modeled as a shell element. The cross section was defined as

five layers of linear elastic orthotropic wood material assuming that the all plastic deformations

would occur in the connectors. Contact springs were also placed at the base of each shell along

the bottom of the wall. The results from this model were compared with the results from wall

tests and were found to match closely in hysteretic behavior as well as total energy dissipated.

Ceccotti (2008) performed an analysis to predict the results of the 3D three story building shake

table test. An analytical model, created in Drain3D, was modified to allow for the type of non-

23

linear behavior of timber connections. The model consists of three major components, rigid

panels modeled as stiff braced frames, and two types of non-linear springs: one type to represent

the angle brackets, with symmetric nonlinear pinching behavior to match experimental data and

the other type to represents the hold-downs, with non-symmetric behavior. Non-linear pinching

behavior is modeled in compression and very stuff linear elastic behavior is modeled in tension.

Dickof (2013) numerically studied CLT-steel hybrid systems at three, six, and nine story heights,

examining the seismic response of this type of hybrid SFRS in regions with moderate to high

seismic hazard indices. A non-linear model of a 2D in-filled frame system was developed and

compared to the behavior of a similar plain steel frame at each height. Parametric analyses were

performed to determine the effect of the panels and the connection configuration, steel frame

design, and panel configuration in a multi-bay system. Static pushover loading was applied

alongside semi-static cyclic loading to allow a basis of comparison to future experimental tests.

Dynamic analyses were run using ten ground motions linearly scaled to the uniform hazard

spectra for Vancouver, Canada with a return period of 2% in 50 years as, 10% in 50 years, and

50% in 50 years to examine the effect of infill panels on the interstory drifts. The ultimate and

yield strength and drift capacity were used to determine the overstrength and ductility factors as

described in the NBCC (NRC, 2010). It was observed that strength and stiffness of the system

increased almost linearly with addition of each CLT panel, while at the same time interstory drift

was reduced. The results showed CLT infill panels are better suited to low ductility systems.

Ductility factor of 3.0 and overstrength factor of 1.3 have been recommended for such system.

24

2.6 FFTT System

2.6.1 Structural System

A new innovative system called FFTT- “Finding Forest through Trees”, predominantly a mass-

timber vertical system bolted with partially embedded steel beam section, has been introduced by

Green and Karsh (2012). No concrete is used beyond grade level and the system relies on steel

sections for ductility. Steel beams have their sections reduced at the desired location to initiate

plastic hinges under seismic loads. Due to the combination of high strength to weight ratio of

mass timber and possible enhancement of lateral strength and ductile behavior due to steel

sections, this system can serve as a viable option for the construction of high-rise timber

structures in future.

The FFTT System consists of large timber panels acting as the vertical system. Beam elements

made of steel sections are bolted to the wall panels and they act as the ductile weak link of the

system (Figure 14). Beams are designed to have reduced cross-section near the end of the beam,

such that plastic hinging occurs at these weak sections at or near design load levels. This

provides the required ductile behavior and resistance to ground shaking.

Figure 14: Solid Panel Core and Intersecting Ductile Steel Link Beams (Green and Karsh, 2012)

25

The four combinations of SFRSs proposed for FFTT System based on the number of stories are

listed in Table 4. The SFRS combination considered in this study is ‘Type 3’, which is a

combination of Structural Core Wall and Perimeter Wall System. The schematic sketch of the

system is shown in Figures 15 and 16. The FFTT system is laterally supported by core wall and

perimeter structural wall. Steel beams run all across the perimeter wall supporting the panels

over the opening and contributing to the overall ductility of the system. The wall is anchored

down using ductile hold downs or dampers and rigid (elastic) shear connectors.

Figure 15: Type 3 Lateral Load Resisting System for FFTT (Green and Karsh, 2012)

Figure 16: Type 3 Lateral Load Resisting System for FFTT (Green and Karsh, 2012)

26

Table 4: FFTT System Options

Option Lateral Load Resisting Combination Storeys

1 Structural Core Wall – Glulam Perimeter Columns 12

2 Structural Core Wall – Interior Shear Walls – Glulam Perimeter Columns 20

3 Structural Core Wall – Perimeter Moment Frame 20

4 Structural Core Wall – Interior Walls and Exterior Moment Frame 30

A good engineering design of hybrid system like FFTT could overcome the challenges faced by

the performance of light frame timber structures and set the standard for the development of

construction technology for safe high-rise timber structures. Further structural analyses, testing

and diligent peer review, however, are necessary to satisfy all code requirements before the

successful implementation of the FFTT system. Advanced dynamic non-linear analyses,

understanding of moment-frame behavior, detailed connections and cost analyses, fire

performance testing, construction and erecting engineering are recommended as future studies.

2.6.2 Experimental Investigations on FFTT Connection

Recently, through collaboration between UBC and FPInnovations, the effect of steel embedment

length on the FFTT connection system was experimentally investigated (Bhat, 2013). The

experimental program included 7 layer CLT panels as primary lateral force resisting system,

connected by steel beams to provide ductility. To investigate the effect of embedment length on

the load deformation response, a total of five different combinations of beam-wall connections

were tested. Three of these lay-outs involved wide flanged section as steel beam while for the

remaining two series, hollow steel sections were used. The CLT panels were 3 m long and 914

mm wide. Among the three tests conducted with wide flange section, one was partially

embedded, next one was fully embedded and the last was also fully embedded but with reduced

27

cross section near the beam wall joint. The embedment length of the beam in each of these

experiments was the total width of the CLT panel (914 mm). Another two series were conducted

using hollow-steel sections with varying embedment lengths. In all five cases, the overhanging

length of the beam was kept constant at 762 mm. The experimental setup is shown in Figure 17.

The test specimens were subjected to quasi-static monotonic and reversed cyclic loading. At six

different locations on the beam (three on the overhanging portion and three inside the CLT

panel), load-deformation responses were obtained.

Figure 17: Typical Setup and Instrumentation (Bhat, 2013)

The set-ups with wide flange beams showed damage to the CLT panel when the moment reached

around 34.3 kN-m. This force produced excessive compressive stress on the CLT panels. Those

set-ups with HSS sections as steel beam reached an ultimate moment of around 13.7 kN-m, and

thereby did not produce stresses to cause noticeable damage to the CLT panels.

28

Cyclic tests showed good hysteretic behavior and a maximum moment of 33.9 kN-m. A typical

load-deformation response from the tests for HSS section is shown in Figure 18.

Figure 18: Load-deformation plot of the test configuration 4 (Bhat, 2013)

Bhat (2013) investigated a total of five different configurations. Even though HSS sections

behaved well, these are not practical for mid-rise building construction. HSS sections are very

small in size and building construction demand significantly larger beam sections. For

construction purpose W sections are preferred. W sections are available at larger sizes to suit the

demand of high-rise buildings. In her tests, she used only one W section size (W 150) and did not

vary the embedment length which could be a very important design parameter. So, it is

imperative to conduct further experiments with W sections incorporating variation in beam size

as well as embedment length and depth to find out if these sections are suitable for the FFTT

system. Also only experimental studies are not adequate to draw conclusion and formulate

design guidelines for a new system. These results must be complimented by numerical studies.

Therefore, further studies (both numerical and experimental) needed to be conducted.

29

Chapter 3: NUMERICAL INVESTIGATION ON FFTT SYSTEM

3.1 Finite Element Model Development

To complement the experimental studies conducted by Bhat (2013), finite-element-analyses

(FEA) were conducted on all test configurations. For this purpose, three-dimensional (3D)

models were developed using the commercial software package ANSYS 14.5 (ANSYS Inc,

2013). The details of modelling assumptions are described in the following.

3.1.1 Modelling of CLT Panels

For modelling of the CLT panels, SOLID186, a higher order 3D, 20-node solid element, was

used that exhibits quadratic displacement behavior. The element is defined by 20 nodes having

three degrees of freedom per node. The element supports plasticity, hyperelasticity, creep, stress

stiffening, large deflection, and large strain capabilities. It also has mixed formulation capability

for simulating deformations of nearly incompressible elastoplastic materials, and fully

incompressible hyperelastic materials. The wood material has been modelled as being a linear

elastic orthotropic material. The mechanical properties of CLT used in the model are shown in

Table 5. The x and y direction properties were altered to represent the different layers of CLT.

The layers of CLT panels are glued together so that force transfer occurs between layers.

Table 5: Properties of CLT

Elastic Moduli(MPa) Poisson Ratio Shear Moduli (MPa)

Ex 11000 vxy 0.40 Gxy 700

Ey 5500 vyz 0.40 Gxy 500

Ez 600 vzx 0.04 Gxy 70

30

3.1.2 Modelling of Steel Beams

Similar to CLT, SOLID 186 elements have been used to model the steel beam. Bilinear isotropic

elasto-plastic material properties have been used to accommodate the post-yield inelastic

response of the steel beam. The material properties are shown in Table 6.

Table 6: Properties of Steel beam

Modulus of Elasticity, E (MPa) 210,000

Yield Strength, fy (MPa) 310

Post-yield Stiffness, α (MPa) 5,000

Ultimate Strength, fu 420

Wide Flange Section W 150 x 26

Hollow Steel Section HSS 100 x 50

3.1.3 Contact Simulation

During the experiments, the steel beam came in contact with the CLT panel as it was pushed. To

simulate this behavior, surface to surface contact technology has been used. This type of contact

provides linear traction-separation, standard contact behavior after debonding and has capability

of modeling unloading and reloading phase. The ANSYS Contact Manager was used to define

areas of contact; the coefficient of friction (μ) between steel and wood has been set to 0.3.

3.1.4 Boundary Conditions

All degrees of freedom were constrained at the base and at the top of the CLT panel. The steel

beam was prevented against lateral buckling by restraining its translation along longitudinal and

lateral direction inside the CLT panel. But the beam was allowed to rotate about its longitudinal

axis. A typical finite element model is shown in Figure 19.

31

Figure 19: Finite Element Model of test configuration 1

3.1.5 Load Application

A concentrated load was applied at the free end of the beam to simulate the actuator load during

the experiments. The load was applied stepwise with small increments of time (0.05 seconds) to

allow the solution to converge.

3.1.6 Post Processing

Upon completion of analysis, results were extracted using the post-processing tool of ANSYS.

The parameters of interest are compressive and shear stress inside the CLT panel, maximum

deformation of the panel, maximum deformation of the steel beam at the free end and load-

deformation behavior. The stress and deformation plots were obtained using “General Post-

processing” feature of ANSYS. The load-deformation curves were constructed by obtaining the

stepwise load and corresponding deformation values using the “Time History Post-processing”

feature of ANSYS.

32

3.2 Numerical Results

3.2.1 Configuration 1: Partially Embedded Wide Flange Section

For this configuration, the wide flange beam was partially embedded inside the CLT panel.

During the experiment, beam yielding occurred at the panel beam interface at average load of 40

kN. The maximum load of 45.8 kN were observed (Bhat, 2013). These values correspond to 30.5

kN-m and 34.9 kN-m bending moment at the same interface, respectively. This configuration

was numerically analyzed and a maximum force of 45.8 kN was applied. The deformation values

were computed at the same six locations as during the experiment. The shear and compressive

stress plots as obtained from ANSYS are shown in figures 20 and 21, respectively. The load-

deformation plots for the cantilever and embedded portions are shown in Figures 22 and 23.

It is observed that the load-deformation curves obtained from the numerical analysis are in good

agreement with the experimental results for both cantilever and embedded portion, thereby

validating the numerical model. However, the degrading portion of the curve was not captured

because of using the bilinear steel material model.

The observed maximum compressive and shear stresses inside the wood were 68 MPa and

14 MPa, respectively. According to the CLT handbook (Gagnon and Pirvu, 2011), the maximum

elastic compressive and shear strength values for CLT are 11.5 MPa and 5.5 MPa, respectively.

Therefore, the observed values have gone well beyond the elastic range indicating that plastic

deformations have occurred. These results also indicate that the use of elastic material model for

CLT is not adequate to obtain actual stress and deformation results. A plastic CLT material

model which considers the post yield behavior of CLT would provide better results.

33

Figure 20: Shear stress plot test configuration 1

Figure 21: Compressive stress plot test configuration 1

34

Figure 22: Comparative load deformation plot of test configuration 1: embedded portion

Figure 23: Comparative load deformation plot of test configuration 1: cantilever portion

0

5

10

15

20

25

30

35

40

0 2 4 6

Moment (kN‐m

)

Deformation (mm)

Loc1_exp

loc1_num

loc2_exp

loc2_num

loc3_exp

loc3_num

0

5

10

15

20

25

30

35

40

0 10 20 30 40 50 60

Moment (kN‐m

)

Deformation (mm)

loc4_exp

loc4_num

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loc5_num

loc6_exp

loc6_num

35

3.2.2 Configuration 2: Fully Embedded Wide Flange Section

For this configuration, as shown in Figure 24, the wide flange beam was fully embedded inside

the CLT panel. During the experiment, beam yielding occurred at the top flange of panel beam

interface at 41 kN and the maximum load was 45.4 kN (Bhat, 2013). These values correspond to

31.25 kN-m and 34.6 kN-m bending moment at the wall beam interface, respectively. Similar to

configuration 1, a force of 45 kN was applied. The deformation values were measured at the

same six locations as during the experiment and compared with. The shear and compressive

stress plots are shown in Figures 25 and 26, respectively. The load-deformation plots for the

cantilever and embedded portions are shown in Figures 27 and 28 and are found to be

reasonable. However, as previously stated, the degrading portion of the curve was not captured

because of using bilinear steel material model. The maximum compressive and shear stress

inside the wood are 83 MPa and 51 MPa, respectively. So again, the observed values have gone

well beyond the elastic range indicating that plastic deformation have occurred.

Figure 24: Finite element model of test configuration 2

36



37



0

5

10

15

20

25

30

35

40

0 2 4 6 8 10

Moment (kN‐m

)

Deformation (mm)

loc1_exp

loc1_num

loc2_exp

loc2_num

Loc3_exp

loc3_num

0

5

10

15

20

25

30

35

40

0 10 20 30 40 50 60

Moment (kN‐m

)

Deformation (mm)

loc4_exp

loc4_num

loc5_exp

loc5_num

loc6_exp

loc6_num

38

3.2.3 Configuration 3: Fully Embedded Section with Reduced Cross Section

The elements of test configuration 3 are shown in Figure 29. It was conducted on fully embedded

wide flange I-sections with reduced cross-section near the beam-panel interface. During the

experiment, beam yielding occurred at the panel beam interface at average load of 44.5 kN

(Bhat, 2013). This value corresponds to 33.9 kN-m bending moment at the wall beam interface.

After numerically analyzing this configuration, the deformation values were computed and

compared with the experimental results. The load-deformation plots for the cantilever and

embedded portions are shown in Figures 30 and 31, respectively. The load-deformation curve

obtained from the numerical analysis is in good agreement with the experimental result for both

cantilever and embedded portion. Reducing the section at the interface did not have significant

effect on the overall behavior of the system. The compressive and shear stress plots show that the

values were lower because of the reduction in steel flange at the panel-beam interface. The

maximum compressive and shear stress inside the wood are 45 MPa and 24 MPa, respectively.

Still, the observed values have gone well beyond the elastic range.

Figure 29: Test configuration 3

39



0

5

10

15

20

25

30

35

40

0 3 6 9 12

Moment (kN‐m

)

Deformation (mm)

Loc1_exp

loc1_num

loc2_exp

loc2_num

loc3_exp

loc3_num

0

5

10

15

20

25

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35

40

0 5 10 15 20 25 30 35 40 45 50

Moment (kN‐m

)

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loc4_exp

loc4_num

loc5_exp

loc5_num

loc6_exp

loc6_num

40

3.2.4 Configuration 4: Full Embedment Length of Hollow Steel Section

For this configuration, hollow structural steel sections were fully embedded inside the CLT

panel. The embedment length was the total width of the panel. During the experiment, beam

yielding occurred at the panel beam interface at 17 kN and the maximum load was 18.5 kN

(Bhat, 2013). These values correspond to 13.0 kN-m and 14.1 kN-m bending moment at the wall

beam interface, respectively. This configuration was numerically analyzed; the finite element

model of this configuration is shown in Figure 26. The deformation values were measured at the

same six locations as during the experiment and compared with. The load-deformation plots for

the cantilever and embedded portions are shown in Figures 33 and 34, respectively. It is

observed that the load-deformation curve obtained from the numerical analysis is in good

agreement with the experimental result for both cantilever and embedded portion. No damage in

the CLT panel was observed during the experiment; from the numerical model it was found that

both horizontal and vertical stress values are too small to cause any damage in the panel.


41



0

5

10

15

20

25

30

35

40

0 2 4 6 8 10

Moment (kN‐m

)

Deformation (mm)

loc1_exp

loc1_num

loc2_exp

loc2_num

Loc3_exp

loc3_num

0

2

4

6

8

10

12

14

16

0 10 20 30 40 50 60 70 80 90 100

Moment (kN‐m

)

Deformation (mm)

loc4_exp

loc4_num

loc5_exp

loc5_num

loc6_exp

loc6_num

42

3.2.5 Configuration 5: Reduced Embedment Length of Hollow Steel Section

For this configuration (Figure 35), hollow structural steel sections were fully embedded inside

the CLT panel. However, the embedment length was reduced to two-third of the width of the

panel. During the experiment, beam yielding occurred at the panel beam interface at 17.1 kN and

the maximum was 18.5 kN (Bhat, 2013). These values correspond to 14.1 kN-m and 14.1 kN-m

bending moment at the wall beam interface, respectively. This configuration was numerically

analyzed and the deformation values were measured at the same six locations as during the

experiment and compared with. The shear and compressive stress plots are shown in Figures 36

and 37, respectively. The load-deformation response for this configuration is shown in Figure 38.

It is observed that the load-deformation curve obtained from the numerical analysis is in good

agreement with the experimental result for both cantilever and embedded portion.


43



44

Figure 38: Comparative load deformation plot of test configuration 5

The compressive and shear stress plots for configurations 4 and 5 are similar. The values were

lower than those corresponding to configurations 1, 2 and 3. The HSS sections had a very small

section modulus compared to the W sections. Therefore, the peak loads for configurations 4 and

5 were much smaller (around 18.5 kN compared to over 40 kN for W sections). The maximum

compressive and shear stress inside the wood for configuration 4 was 22 MPa and 9 MPa,

respectively; while these values were 24 MPa and 10 MPa for configuration 5. Still, the observed

values have gone well beyond the elastic range indicating that the use of an linear-elastic

material model for CLT is not adequate to obtain actual stress results. A non-linear CLT material

model which considers the post yield behavior of CLT would provide better results.

0

2

4

6

8

10

12

14

16

0 10 20 30 40 50 60 70 80 90 100

Moment (kN‐m

)

Deformation (mm)

Loc1_exploc1_numloc2_exploc2_numloc3_exploc4_exploc4_numloc5_exploc5_num

45

3.2.6 Summary on Model Results from Previous Tests

A summary of stress and deformation results for each configuration is presented in Table 7.

Table 7: Results from previous experimental tests and numerical simulation

Series Maximum deformation inside CLT at peak load (mm)

Stresses from numerical analysis (MPa)

Experimental (Bhat, 2013)

Numerical Compressive Shear

Configuration 1 4.0 4.0 68.4 13.2

Configuration 2 7.5 7.0 83.7 21.1

Configuration 3 6.0 5.5 45.3 16.1

Configuration 4 0.55 0.25 24.1 11.6

Configuration 5 5.0 7.0 28.3 13.5

Overall, it is observed that the load-deformation curves obtained from the numerical analyses are

in close agreement with those extracted from experimental investigations. However, the stress

plots from the numerical analyses show that both compressive and shear stresses for each

configuration were beyond the elastic strength limit of timber. Therefore, the connections have

undergone plastic deformation and the linear elastic material model for CLT is no longer

adequate to evaluate the stress magnitudes. Nonlinear material models for CLT have to be

developed to obtain more realistic stress results.

3.3 Numerical Study to Improve Connection Configuration

The experimental tests conducted by Bhat (2013) were significant, since they marked the

beginning of research on the FFTT system at the component level. The results indicated that the

HSS section allowed the desired failure mechanism to form, while the wide flange section

46

caused damage to the CLT panel. However, the hollow section was very small and, therefore, not

suitable for high-rise construction. Moreover, in the tests conducted by Bhat (2013), the beams

were not properly restrained against buckling, which is a critical consideration. Considering

these shortcomings, an attempt has been made to improve the connection configuration. Only

wide flange sections were considered since bigger sized sections can be used.

3.3.1 Geometry

The CLT panels considered were similar to those used for testing by Bhat (2013). ASTM A992

WF beams with 350 MPa yield strength were chosen, additionally, bearing plates and side plates

of the same steel property were included. A total of four bearing plates (each 150 mm in length

and 6.25 mm thick) were placed at top and bottom of the beam to avoid stress concentration at

the face of the panel beam interface. Moreover, to prevent buckling, four side plates of 6.25 mm

thickness were also placed along the web of the beam. Also, the beam was supported against

lateral movement. The material properties are the same as those shown in Tables 4 and 5. The

model is shown in Figure 39. In Figure 40, the beam with the bearing and side plates is shown in

detail.

47

Figure 39: Finite element model for numerical optimization 1

Figure 40: Details of the steel beam with bearing and side plates

48

3.3.2 Parameter Variation

For the purpose of the numerical study; the dimensions of the CLT panel and side plates were

kept constant. The parameters which were varied include spacing between bearing plates,

embedment length and steel beam size and the bearing plate length. The ranges within which

these parameters were varied are shown in Table 8.

Table 8: Parameter range for numerical study

Parameter Range

Embedment length (mm) 500, 600, 700, 800, 900

Spacing between bearing plates ( mm) 250, 300, 350, 400, 450

Steel beam W100 x 19.3, W130 x 23.8, W150 x 29.8

Bearing plate length (mm) 100, 125, 150

3.3.3 Parametric Study Results

A number of analyses have been carried out to observe the behavior of the connection by varying

the parameters as shown in Table 8. The maximum compressive and shear stresses parallel to

grain and deformations inside the CLT panel for various combination of parameters are

summarized in Tables 9, 10 and 11.

49

Table 9: Results of parametric study (Beam: W 150 x 29.8)

Parameters Results

Embedment length

(mm)

Plate length

(mm)

Plate Spacing

(mm)

Compressive stress*

(MPa)

Shear stress*

(MPa)

Deformation inside CLT

(mm)

500

100 400 275 145 62

125 375 267 142 59

150 350 259 141 56

600

100 500 234 136 58

125 475 229 136 56

150 450 217 132 52

700

100 600 202 129 55

125 575 198 127 51

150 550 197 124 47

800

100 700 179 124 48

125 675 173 121 43

150 650 169 120 41

900

100 800 154 119 44

125 775 146 117 40

150 750 141 115 35

*Parallel to grain

50


Parameters Results

Embedment length

(mm)

Plate length

(mm)

Plate Spacing

(mm)

Compressive stress*

(MPa)

Shear stress*

(MPa)


(mm)

500

100 400 243 134 46

125 375 237 131 45

150 350 231 128 43

600

100 500 217 130 47

125 475 212 126 43

150 450 207 125 42

700

100 600 197 128 45

125 575 188 125 41

150 550 179 121 41

800

100 700 165 115 40

125 675 157 112 37

150 650 151 110 33

900

100 800 131 106 34

125 775 121 101 30

150 750 119 97 26

*Parallel to grain

51


Parameters Results

Embedment length

(mm)

Plate length

(mm)

Plate Spacing

(mm)

Compressive stress*

(MPa)

Shear stress*

(MPa)


(mm)

500

100 400 195 107 33

125 375 191 104 31

150 350 186 103 29

600

100 500 165 101 30

125 475 161 96 27

150 450 156 95 26

700

100 600 137 90 28

125 575 132 86 25

150 550 129 83 23

800

100 700 112 81 24

125 675 106 74 22

150 650 104 72 21

900

100 800 91 73 20

125 775 84 65 17

150 750 79 59 16

*Parallel to grain

52

It is observed that the parallel to grain compressive stress values remain very high and indicate

crushing. A reduction in stress values has been achieved by increasing the embedment length and

spacing between bearing plates. However, stresses still remain over the elastic limit. It means

that for the given panel dimension and strength property, larger wide flange sections might be

too strong and will cause crushing in the panel. However, this conclusion is drawn based on the

elastic material model of CLT which is not sufficient to represent the actual behavior.

Upon reaching this conclusion, the smallest commercially available Wide Flange section (W 100

x 19) was chosen for the further analyses. To extract the load-deformation behavior, four points

were chosen marked as LVDT-1 and LVDT-2 (inside the CLT panel) and LVDT-3 and LVDT-4

(in the beam at the cantilever portion). During the subsequent experiments (as reported in chapter

4), these locations were used to instrument the test specimens with Linear Variable Differential

Transformers (LVDTs) and compare the experimental results to the numerical results.

In the model, the system yielded at an applied moment of 30.2 kN-m at the beam-wall interface.

The maximum deformation computed at the end of the beam was 15 mm. The connection

continued to pick up load up to 42.1 kN-m. The deformations inside the CLT panel were small as

can be seen from the load-deformation plot (Figure 41). Observing the sign of displacement

values, it can be concluded that the beam rotated about a point between LVDT- 1and LVDT-2.

The compressive stresses are still very high (in the region of 80 MPa). The compressive stress

plot is shown in Figure 42. However, this large stress occurred only within a very small region at

the beam-panel interface. Therefore, unless the CLT panel dimension and strength properties are

increased; only small Wide-Flange sections might provide the expected ductile behavior. Hence,

to experimentally validate this outcome, W100 x19 section was chosen for the subsequent

experimental investigations.

53

Figure 41: Load-deformation plot at different points of interest for the optimization study when the

W100 x19.3 beam was fully embedded with 150 mm bearing length and 350 mm spacing

Figure 42: Contour plot of compressive stress parallel to grain inside the CLT panel when the

W100x19.3 beam was fully embedded with 150 mm bearing length and 350 mm spacing

0

5

10

15

20

25

30

35

40

45

‐5 5 15 25 35 45 55 65

Moment (kN‐m

)

Deformation (mm)

LVDT‐1

LVDT‐2

LVDT‐3

LVDT‐4

54

3.3.4 Parametric Study with Partial Embedment Depth

The previous results have been obtained when the embedment depth is the full depth of the

beam. During experimental testing of this configuration, rolling shear failure occurred as

explained in Chapter 4. CLT is very weak against rolling shear; to avoid such failure, another

numerical analysis has been carried out with a partial embedment depth of the steel beams. The

embedment depth considered was 85 mm instead of 102 mm. Also the distance between bearing

plates and consequently the embedment length were increased to reduce the bearing force. For

this analysis the center to center distance between bearing plates was increased to 665 mm (from

350 mm). The model is shown in Figure 43.

Figure 43: Finite element model for numerical optimization 2

55

The compressive stress plot for this improved configuration is shown is Figure 44. The results

show significantly reduced compressive and shear stress values. The compressive stress (parallel

to grain) reduced from 186 MPa to 102 MPa while the shear stress (parallel to grain) dropped to

63 MPa from 76 MPa. This is due to the fact that the bearing force has been reduced. A ductile

failure mode is predicted with large deformation in the steel beam. The free end of the beam

deformed up to 125 mm whereas the deformation inside the CLT panel is around 2 mm. The

load-deformation plot for this improved configuration is shown in Figure 45.

Figure 44: Contour plot of compressive stress parallel to grain numerical model when the W100

x19.3 beam was partially embedded with 150 mm bearing length and 665 mm spacing

56

Figure 45: Load-deformation plot at different points of interest for the optimization study when the

beam was W100 x19.3 with 150 mm bearing length and 665 mm spacing

3.4 Discussion on Numerical Analysis and Optimization Studies

The numerical studies complemented the experimental investigation by Bhat (2013). It

demonstrated that the HSS section is well suited to FFTT connection system. The HSS section

being very small, caused small compressive and shear stress to the CLT. It also failed in a ductile

manner with large deformation which is desirable for the connection. The HSS section, however,

is not practical for the actual construction of structures owing to its very small size. From a

practical point of view, wide flange sections are better suited since larger sections are available.

However, when wide flange sections were used for the study, the wood was subjected to

excessive stress which may lead to failure before steel. The obtained compressive and shear

stresses were beyond the elastic strength limit specified by the CLT Handbook (Gagnon and

Pirvu, 2011).

0

5

10

15

20

25

30

35

40

‐5 20 45 70 95 120

Moment (kN‐m

)

Deformation (mm)

LVDT‐4

LVDT‐3

LVDT‐2

LVDT‐1

57

To search for a better connection configuration, a numerical parameter study was carried out.

The study revealed that by incorporating bearing and side plates the connection behavior can be

significantly improved. As can be seen from Tables 9, 10 and 11 all the parameters of interest

impact the amount of stress and deformation magnitudes inside the CLT panel.

The size of the beam is a major factor. The greater the beam size, the more stress and

deformation it causes to the beam which is expected. During this parametric study, three beams

were chosen with the largest being W 150x 29.8 and smallest being W100 x19.3, while the other

section being W 130x 23.8. The difference in elastic section modulus between these beams is

quite significant. W 150x29.8 has elastic section modulus of 218.4 cm3. The corresponding

elastic section moduli of W 130x 23.8 and W100 x 19.3 are 139.5 cm3 and 89.9 cm3 respectively.

The W 150x 29.8 section caused much higher compressive and shear stresses as well as larger

deformations inside the CLT panel than the smaller sections given that other parameters remain

same. For the W150 x29.8 section, the maximum parallel to grain compressive and shear stresses

observed were 275 MPa and 145 MPa, respectively. This was observed when 100 mm bearing

plates at 400 mm spacing was considered. It is also noticeable that by increasing the bearing

plate length to 150 mm and spacing to 775 mm, the compressive stress could be reduced to 141

MPa (49% decrease). The shear stress also reduced from 145 MPa to 115 MPa (21% decrease).

For the W130 x23.8 section, the maximum compressive and shear stresses observed were 243

MPa and 134 MPa, respectively. These stresses are much lower than those caused by the W 150

x 29.8 section. This was observed when 100 mm bearing plates at 400 mm spacing was

considered. By increasing the bearing plate length to 150 mm and spacing between them to 775

mm, the compressive stress could be reduced to 119 MPa (49% of maximum value). The

reduction in shear stress was 20 % (from 134 MPa to 107 MPa).

58

For the W100 x19.3 section, which is the smallest commercially available wide flange section,

the maximum compressive and shear stresses observed were 195 MPa and 107 MPa,

respectively. Similar to the behavior observed during optimization with bigger sections, the

highest stress values were observed when the lengths of the bearing plates were the shortest (100

mm) and the spacing between the plates was 400 mm. Again, by increasing the bearing plate

length to 150 mm and spacing between them to 775 mm, the compressive stress could be

reduced to almost 41% (from 195 MPa to 79 MPa). The shear stress also reduced significantly

(from 107 MPa to 59 MPa).

Form Tables 9, 10 and 11 and the preceding discussion, it is obvious that the length of the

bearing plates, the embedment length of beam inside CLT and, consequently, the spacing

between bearing plates are critical factors for connection design optimization. The effect of

increasing the embedment length and therefore spacing between bearing plates are very

significant as can be seen in Figure 46. This figure shows the variation in compressive stress

with increasing embedment length of beam for a W 100x 19.3 section. Significant reduction in

compressive stress was achieved by bearing plate length same but increasing embedment length.

However, the variation in compressive stress with bearing plate length is not as pronounced

which can be seen from Figure 47. This figure shows the variation in compressive stress with

bearing plate length for the W100 x 19.3 beam. The reduction in stress when just increasing the

bearing plate length is negligible.

Similar to the compressive stress, the shear stress can also be significantly reduced by increasing

the spacing between bearing plates and the length of the plates. Again, the effect of increased

embedment length and spacing between plates is more pronounced (Figure 48), while the

influence of bearing length is negligible (Figure 49). As the stresses are reduced significantly, so

59

do the displacement values with increasing spacing between plates and the length of the plates.

The variation of displacements with embedment length and with bearing plate length are shown

in Figures 50 and 51 for the W100 x 19.3 beam.

When the spacing between bearings plates are increased, the lever arm for resisting the applied

moment increased, therefore, the forces transferred through the bearing plates are significantly

reduced. The effect of increased bearing length means greater bearing area for force transfer

between beam and CLT panel. However, when the length of plate is increased, the center to

center spacing between them is decreased if the embedment length of the beam is kept constant.

This might be the cause of not significant reduction in stresses even though the bearing area is

increased.

For the preceding discussion, the limitation of the linear-elastic material model for CLT has to be

kept in mind: all the stress values are purely numerical results and can only be used for

comparative purposes. In reality, the localized stress peaks would lead to wood crushing and

plastic deformation with a resulting stress redistribution over a larger area but with a

significantly reduces stress magnitude.

Overall, on the basis of the linear elastic CLT material model, it was found that even the smallest

wide flange steel section was stronger than the CLT panel. While it implies that wide flange

section is stronger than the CLT and would not provide ductile failure mode; it also indicates that

the linear elastic model for CLT used for the numerical study is not entirely sufficient. Nonlinear

material model for CLT will better simulate the actual behavior of the system.

60

Figure 46: Variation in compressive stress with embedment length of beam (Beam: W 100 x 19.3)

Figure 47: Variation in compressive stress with length of bearing plate (Beam: W 100 x 19.3)

60

80

100

120

140

160

180

200

500 600 700 800 900

Compressive stress (Mpa)

Embedment length (mm)

100 mm bearing length



60

80

100

120

140

160

180

200

100 125 150

Compressive stress (Mpa)

Bearing plate length (mm)

500 mm embedment length





61

Figure 48: Variation in shear stress with embedment length of beam (Beam: W 100 x 19.3)

Figure 49: Variation in shear stress with length of bearing plate (Beam: W 100 x 19.3)

50

60

70

80

90

100

110

500 600 700 800 900

Shear stress (Mpa)





50

60

70

80

90

100

110

120

100 125 150

Shear stress (Mpa)







62

Figure 50: Variation in displacement with embedment length of beam (Beam: W 100 x 19.3)

Figure 51: Variation in displacement with length of bearing plate (Beam: W 100 x 19.3)

10

15

20

25

30

35

40

500 600 700 800 900

Deform

ation inside CLT (mm)





10

15

20

25

30

35

40

100 125 150

Deform

ation inside CLT (mm)







63

Chapter 4: EXPERIMENTAL INVESTIGATION ON FFTT SYSTEM

4.1 Introduction

This chapter describes the experimental tests conducted on the improved connection

configuration obtained by the numerical study described in Chapter 3. The tests evaluated the

behavior of embedded wide flange sections through quasi-static monotonic and reverse cyclic

tests. Component tests with two different configurations were conducted in the Structural

Laboratory of FPInnovations, Vancouver. The objective of the experimental study was to

observe if the new connection layouts initiate the desired “Strong column-week beam” failure

mode.

4.2 Materials

Two 7-ply CLT panels of grade S-P-F No.1/No.2 of 0.9 m wide and 4 m long were used. The

outer laminations were 32 mm while the inner laminations were 35 mm thick because the

surfaces were planned. The overall thickness of the panel was 239 mm. The design material

properties listed by the manufacturer of the CLT product (Structurlam) used in the project are

summarized in Table 8.

ASTM A992, W 100 x 19 sections were chosen. The yield strength and ultimate strength of the

steel specimens were 350 MPa and 460 MPa, respectively. The modulus of elasticity was taken

to be 210 GPa. As bearing and side plates, rectangular flat steel bars of 150 x 100 x 6.25 and 87

x 50 x 6.25 were used, respectively.

64

4.3 Specimen Description

The slots, into which the beam sections were embedded, were pre-cut in the CLT panel. A total

of 3 slots were cut in each CLT panel to facilitate two static and one cyclic test. The beams

embedded into these slots were held in place using two 9.5 mm lag bolts in 12.7 mm drill holes,

at 250 mm and 457 mm from wall beam interface for series 1 and 2, respectively. The

experimental setup for series 1 along with the position of the four LVDTs (red arrows) is shown

in Figure 52. In Figure 53, a side view of the embedment of the beam is shown.The force transfer

through bearing of bolts was assumed to be negligible. Complete load transfer occurred through

the bearing of steel beams alone. Two series of tests were conducted with two replicates of

monotonic test and one cyclic test for each series as shown in Table 12.

Table 12: Test specimen description

Series Embedment Embedment Length Bolted Connection Distance between plates

1 102 mm 500 mm 9.5 mm diameter bolt at 250 mm from interface

350 mm c/c

2 85 mm 914.4 9.5 mm diameter bolt at 457 mm from the

interface

664.4 mm c/c

65

Figure 52: Experimental setup for test series 1

Figure 53: Full embedment of the steel beam inside the CLT panel for test series 1

Series 2 was conducted by embedding the wide-flange I-section 85 mm into the outer 3 plies of

the panel. The experimental setup for the partially embedded beams in series 2 is shown in

66

Figure 54. Series 2 was conducted by embedding the section 85 mm into the outer three plies in

order to avoid rolling shear failure and to observe if avoiding the rolling shear failure can

improve the behavior of the system.

Figure 54: Full embedment of the steel beam inside the CLT panel for test series 2

4.4 Test Procedure

The panels were bolted down to the floor at both ends to restrain them from translation, rotation

or uplift movement during the experiments. For series 1, four LVDTs were attached to the

embedded beam with the first LVDT placed 152.4 mm from the edge of the panel (as shown by

red arrows in Figure 52). For series 2, similarly four LVDT s were attached to the beam. The two

LVDTs placed inside the CLT panel were located at the center of bearing plates. The LVDTs

placed at the cantilever portion were located at 350 mm and 700 mm away from the beam-wall

interface.

In the quasi static monotonic tests, the load was applied at the end of the projecting beam

through a calibrated actuator (225 kN capacity). The loading was maintained constant at a rate of

12.7 mm/min. For series 1, the load kept on increasing without dropping. Hence no peak load

67

was reached and the loading for the monotonic tests was discontinued when the applied load

reached 62 kN, which caused 43.4 kN-m at the beam-wall interface. However, the system

showed well defined yield point at 52 kN force (36.4 kN-m). For series 2, similar behavior was

observed with well-defined yield point at 45.0 kN force (32.6 kN-m) and the load continued to

increase to 56 kN (40.6 kN-m) without degrading. The deformations at 90% yield load from the

monotonic tests were chosen as target displacements (100%) for the subsequent reversed cyclic

loading tests.

The CUREE protocol (Krawinkler et al., 2001) was used for the cyclic loading for each test

series (Figure 55 and 56). The loading was programmed to continue with an increment of 20 %

beyond the target displacement until 200% of the target displacement. The cyclic tests were

conducted at a loading rate of 5 mm/min (equivalent to a rotation of the beam of 0.007 rad/min).

Figure 55: CUREE loading protocol for series 1

‐180‐160‐140‐120‐100‐80‐60‐40‐200

20406080100120140160180

0 100 200 300 400 500 600 700 800

Displacement (%

of Max)

Time (seconds)

Cyclic Displacement ScheduleCUREE Test Protocol

100% displacement = 1.125"Load rate= 0.2"/min

68

Figure 56: CUREE loading protocol for series 2

4.5 Experimental Results

4.5.1 Series 1: Monotonic Test on Fully Embedded Beam

During the tests, beam yielding occurred at the panel-beam interface (Figure 57) at an average

35.1 kN-m bending moment. Both tests showed well defined yield points. However the

deformations were larger in test 1 compared to test 2. The system kept on taking load without

dropping and no peak load was identifiable. Both tests were terminated when the moment

reached 42.8 kN-m at the beam-wall interface.

-180-160-140-120-100

-80-60-40-20

020406080

100120140160180

0 200 400 600 800 1000 1200 1400

Dis

pla

cem

ent

(% o

f M

ax)

Time (seconds)

Cyclic Displacement ScheduleCUREE Test Protocol

100% displacement = 2.000"Load rate= 0.2"/min

69

Figure 57: Yielding of beam during experimental series 1

Compared to the load-displacement curve obtained from numerical study (Figure 41), these

monotonic test results are in close agreement. The curve shapes are similar. But the experimental

yield point (36.4 kN-m moment) was 20% higher than numerical value (30.1 kN-m moment).

Also the yield displacement at the location of LVDT 4 is similar to numerically obtained value

(17 and 15 mm). The deformation that occurred inside the CLT panel during monotonic test of

series is shown in Figure 58.

70

Figure 58: Deformation inside the CLT panel during experimental series 1

During both tests, the maximum deformation at the location of LVDT- 4 was around 60 mm

which occurred at the 42.8 kN-m moment. At the location of LVDT-3, the deformation at yield

moment was around 6 mm, whereas the maximum deformation at highest moment was 25 mm.

The difference between yield and maximum displacement indicates that the system has good

ductility. The load-displacement plots of the monotonic tests 1 and 2 of series 1 are shown in

Figures 59 and 60 respectively.

71

Figure 59: Load-displacement curve: Series-1, monotonic test-1


0

5

10

15

20

25

30

35

40

45

50

‐5 5 15 25 35 45 55 65

Moment (kN‐m

)

Deformation (mm)

LVDT‐1

LVDT‐2

LVDT‐3

LVDT‐4

0

5

10

15

20

25

30

35

40

45

50

‐5 5 15 25 35 45 55 65

Moment (kN‐m

)

Deformation (mm)

LVDT‐1

LVDT‐2

LVDT‐3

LVDT‐4

72

The in-plane deformation of the embedded portion of the beam was measured at two locations.

LVDT-1 corresponds to the LVDT attached at the far end of the embedded beam section. The

deformation values obtained from both the LVDTs inside the CLT panel were very small

(around 2 mm). The negative displacement value of LVDT-1 and positive displacement values of

LVDT-2 indicate that the beam was rotated about a point between these two. The data acquired

at LVDT-1 during test 1 was erroneous due to slot fabrication imperfections (e.g. slight spaces

between beam and wood) that existed prior to the testing. This error was avoided during test 2 by

improving the quality of fabrication. The stiffness of both curves were similar.

There were in-plane deformations inside the panel causing damage to wood before the beam

reached yield load. Even though the damage was negligible, it still indicated that the beam is

stronger than the CLT panel. The beam chosen, being the smallest commercially available

section, reinforces the fact that this connection layout is not ideal for the FFTT system. Further

studies need to be conducted with improved connection configuration and nonlinear plastic

material properties for CLT, before this system can be considered.

4.5.2 Series 1: Cyclic Test on Fully Embedded Beam

No peak load was identifiable from the monotonic tests, but yield point was well defined.

Therefore, the deformation at 90% of the peak load was considered as target displacement. The

value of this displacement was 28 mm. The setup was similar to monotonic tests. However, the

failure mode was rolling shear followed by crushing. The beam rotated about the point where

bolt was inserted. Rolling shear crack was seen at the weaker layer of the CLT panel as shown in

Figure 61. Cracking and crushing began inside the panel before the beam began to yield. The

load was continued to reach 180% of the target displacement.

73

The observed maximum bending moment at the beam-wall interface was 50.1 kN-m. The

hysteresis behavior of the top face flange at the location of LVDTs is presented in Figures 62

through to 64. Based on the monotonic and cyclic tests, it can be deduced that the point of

rotation of the beam is between LVDT-1 and LVDT-2. The maximum deformation at LVDT-1

which is located 425 mm inside from the beam-wall interface was found to be very close zero

(3.8 mm), with negligible energy dissipation. The maximum deformation at LVDT-2 which is

located 75 mm inside from the beam-wall interface was also found to be very small (4.4 mm),

with negligible energy dissipation.

Figure 61: Rolling shear failure in CLT panel during cyclic test of series 1

74

The hysteretic curves obtained from LVDT-1 and 2 (Figures 62 and 63) show a stiffer slope on

one side and flatter slope on the opposite. Very little energy dissipation occurred inside the CLT

panel. The readings from LVDT-3, located 375 mm away at the cantilever portion of the beam,

were erroneous and therefore not shown. Hysteresis plots at the locations of LVDT- 4 (Figure

64) suggest that almost all energy under cyclic loading was dissipated through the deformation of

the cantilever portion of the beam for which the maximum deformation was 48 mm. The cyclic

test, like the monotonic test showed that damage occurred in the CLT panel in the form of rolling

shear before the steel beam yielded.

Figure 62: Cyclic test: Series-1, LVDT-1

‐60

‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐4 ‐3 ‐2 ‐1 0 1

Moment (kN‐m

)

Deformation (mm)

LVDT‐1

75



‐60

‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐5 ‐4 ‐3 ‐2 ‐1 0 1 2

Moment (kN‐m

)

Deformation (mm)

LVDT‐2

‐60

‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐60 ‐50 ‐40 ‐30 ‐20 ‐10 0 10 20 30 40

Moment (kN‐m

)

Deformation (mm)

LVDT‐4

76

4.5.3 Series 2: Monotonic Test on Partially Embedded Beam

During the tests, beam yielding occurred at the panel-beam interface (Figure 65) at an average

33.4 kN-m bending moment. Both the tests showed well defined yield points. The system kept

on taking load without dropping and no peak load was identifiable. Both the tests were

terminated when the moment reached 38.5 kN-m at the beam-wall interface. Compared to series

1, the yield and maximum load obtained at series 2 are slightly lower. The longer distance

between the bearing plates and consequently the greater lever arm for resisting moment

contributed to the system yielding at a lower moment.

Figure 65: Yielding of beam during experimental series 1

77

Comparing the experimental curve to the load-displacement curve obtained from numerical

optimization (Figure 45), these monotonic test results are in close agreement. The curve shapes

are similar. But the experimental yield point was slightly higher than numerical value. This

discrepancy indicates that the actual system might be stiffer than the numerical model. Also the

yield displacement at the location of LVDT-4 is very similar to numerically obtained value (26

and 29 mm respectively).

During test 1, the maximum deformation value of at the location of LVDT- 4 was around

180 mm which occurred at 38.5 kN-m moment. From test 2, the deformation at the same location

was 140 mm. At the location of LVDT-3, during first monotonic test, the deformation at yield

moment was around 12 mm, whereas the maximum deformation at highest moment was 75 mm.

The large difference between yield and maximum displacement indicates that, the system has

greater ductility than that of series 1. The load-displacement plots of the monotonic test 1 and 2

of series 1 are shown in Figures 66 and 67 respectively.


0

5

10

15

20

25

30

35

40

45

‐5 15 35 55 75 95 115 135 155 175 195

Moment (kN‐m

)

Deformation (mm)

LVDT‐1

LVDT‐2

LVDT‐3

LVDT‐4

78


The in-plane deformation of the embedded portion of the beam was measured at two locations.

The deformation values obtained from both the LVDTs inside the CLT panel were very small

(around 2 mm). The negative displacement value of LVDT-1 and positive displacement values of

LVDT-2 indicate that the beam was rotated about a point between these two. No out of plane

buckling was observed during monotonic testing of series 2 configuration. There was in-plane

deformation inside the panel causing damage to wood before the beam reached yield load. Even

though the damage was negligible, it still indicated that the beam is stronger than the CLT panel.

4.5.4 Series 2: Cyclic Test on Fully Embedded Beam

Similar to monotonic test of series 1, no peak load was identifiable from the monotonic tests for

series 2. However, considering greater ductile behavior of this system as demonstrated by

monotonic tests, a higher deformation (corresponding to 100 % yield load) was considered as

target displacement. The value of this displacement was 50 mm. The setup was similar to

monotonic tests. However, the failure mode was yielding of the beam followed by out of plane

0

5

10

15

20

25

30

35

40

45

‐5 15 35 55 75 95 115 135 155

Moment (kN‐m

)

Deformation (mm)

LVDT‐1LVDT‐2LVDT‐3LVDT‐4

79

buckling. No rolling shear failure was observed during cyclic test of series 2 which is a big

improvement towards search for ideal connection configuration. The beam rotated about the

point where bolt was inserted. The out of plane buckling failure mode of the beam during cyclic

test are shown in Figure 68. Cracking and crushing began inside the panel when the load was

beyond the yield load of the system. It was planned to continue the load up to 200% of target

displacement. However at 160% of the target displacement, out of plane buckling occurred in the

beam and it lifted up from its position. At this point, the application of load was discontinued.

Figure 68: Out of plane buckling of the steel beam during cyclic test of series 2

Minor damage in the form of splitting of CLT at the location of bearing plate was also observed

as shown in Figure 69. The observed maximum bending moment at the beam-wall interface was

38.5 kN-m. The hysteresis behavior of the top face flange at the location of LVDTs is presented

in Figures 70 through to 73. Based on the monotonic and cyclic tests, it can be deduced that the

80

point of rotation of the beam is between LVDT- 1 and LVDT-2. The maximum deformation at

LVDT-1 which is located 675 mm inside from the beam-wall interface was found to be around

10 mm with small energy dissipation. The maximum deformation at LVDT-2 which is located

125 mm inside from the beam-wall interface was also found to be small (8 mm), with little

energy dissipation. Upon completion of cyclic test a plastic deformation of 5 mm was observed

at the CLT layer in contact with the bearing plate.

Overall, series 2 showed ductile behavior with steel beam yielding before any significant damage

to CLT. So, in terms of performance, the partially embedded system is better compared to the

fully embedded system. However, the beam used during the experiment was still the smallest one

commercially available. Therefore, further testing is required with larger beam size before it can

be concluded that partially embedded system is an ideal configuration for the FFTT system.

Figure 69: Damage in the CLT panel during cyclic test of series 2

81



‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐12 ‐10 ‐8 ‐6 ‐4 ‐2 0 2 4 6

Moment (kN‐m

)

Deformation (mm)

LVDT‐1

‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐10 ‐8 ‐6 ‐4 ‐2 0 2 4

Moment (kN‐m

)

Deformation (mm)

LVDT‐2

82



‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐40 ‐30 ‐20 ‐10 0 10 20 30 40

Moment (kN‐m

)

Deformation (mm)

LVDT‐3

‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐80 ‐60 ‐40 ‐20 0 20 40 60

Momen

t (kN‐m

)

Deformation (mm)

LVDT‐4

83

4.6 Discussion on Experimental Investigations

To substantiate the findings from the numerical parameter study, two experimental test series

were conducted. Both series included two monotonic and one reversed cyclic test. The

experimental results are discussed in the subsequent paragraphs.

4.6.1 Comparison between Experimental and Numerical Results

A comparative summary of experimental and numerical results is shown in Table 13. It can be

observed that for both the series, the numerical yield moments are around 12% lower than those

obtained from experiments. However, the peak moments are almost identical. Therefore, the

numerical model seems appropriate although a little less stiff than the actual connection. It is also

noticeable from both experimental and numerical studies that a partially embedded beam with

greater embedment length (series 2) yielded at a lower load than the system with full embedment

of beam with reduced embedment length (series 1). The longer embedment length of beam inside

CLT resulted in a longer lever arm for resisting the external force. Also, partial embedment

caused the beam to lift up from its position due to out of plane buckling, therefore might be a

limiting factor.

Table 13: Comparison between Experimental results and their numerical simulation

Series

Yield Moment

(kN-m)

Peak Moment

(kN-m)

Maximum Deformation at LVDTs

(mm)

Exp. Num. Exp. Num. 1 2 3 4

Exp. Num. Exp. Num. Exp. Num. Exp. Num.

1 36.4 30.1 42.8 42.4 -3.2 -2.2 1.4 1.8 34.2 30.2 60.4 68.0

2 33.4 29.5 38.5 38.4 -1.4 -1.2 0.8 0.9 74.4 55.0 175.0 130.0

84

The 1st and 2nd LVDTs were placed inside the CLT and the other two at the cantilever portion of

the beam. The experimental deformation values inside the CLT for series 1 are greater than those

obtained from series 2. This can be explained by the greater lever arm for the resisting moment

in case of series 2 causing smaller compressive forces inside the CLT and consequently smaller

deformations. This fact is supported by the numerical simulation which showed lower

deformation inside CLT for series 2. The other two LVDTs represent the deformation of the steel

beam only. For these two locations, series 2 showed lower deformations than series 1. This can

be attributed to the fact that the fully embedded beam resulted in a stiffer system with no out of

plane buckling, while partial embedment of beam in series 2 caused the beam to buckle out of

plane resulting in significantly greater deformation of beam. The numerical analyses showed

similar behavior but the values were lower than the corresponding experimental results. This is

because the numerical model considered steel as a bilinear material without degradation while

the actual steel exhibits degrading behavior. Overall though, the experimental and numerical

results for the monotonic tests are in reasonable agreement.

4.6.2 Point of Rotation of Beam

The point of yielding was at the beam-wall interface for both the test series. The point of beam

rotation inside the CLT however varied between series. The point of rotation of beam was

established based on the load-deformation behavior of the LVDTs attached inside CLT (LVDT-1

and LVDT-2). The point of rotation of beam is the location at which the deformation and

consequently energy dissipation is zero. By observing the signs of displacement values inside

CLT from Table 13, it can be concluded that the points of rotation of beam for both the systems

lie between LVDT-1 and LVDT-2. For both series, deformation values from LVDT-1 are

negative and the values from LVDT-2 are positive. A closer inspection of the values also

85

revealed that the point of rotation is closer to LVDT-2 for both systems. In Figure 74, the

displacement values of LVDT-1 and 2 are plotted against the location of LVDTs. The value 1

and 2 in the horizontal axis depicts the position of LVDT-1 and 2 respectively. By assuming

linear variation of displacement, it can be showed that, for series 1, axis of the beam rotation is at

1.7 times the distance from LVDT-1 towards LVDT-2. For series 2 this location is at 1.6 times

the distance between two LVDTs from LVDT-1 towards LVDT-2. Such assumption is

reasonable considering very small displacement values at these locations. The points of rotation

are illustrated in Figure 74. These points are of interest for the calculation of the stress transfer

between steel and CLT through the bearing of the embedded beams on the wall panels.

Therefore, these points are important for obtaining the location of bearing plates to achieve

optimal performance of the connection.

Figure 74: Points of rotation of beams for series 1 and 2

4.6.3 Ductility and Force Modification Factor

Ductility (µ) can be calculated from the load-displacement curves from the monotonic tests. It is

an important parameter for seismic design. The ratio of ultimate and yield displacement can be

‐4

‐3

‐2

‐1

0

1

2

1 1.1 1.2 1.3 1.4 1.5 1.6 1.7 1.8 1.9 2

Deform

ation (mm)

LVDT 1 and 2 location

Series1

Series2

86

considered as ductility for the connection (Munoz et al., 2009). The ductility related force

modification factor (Rd) specified in NBCC 2010 (NRC, 2010) can be obtained from the

connection ductility (Boudreault et al., 2007):

…………………………………………………………………………… (4)

Table 14 shows the values obtained from the experiments and applying the above equation.

Table 14: Ductility ratio and force modification factor

Series Ductility ratio (µ) Force modification factor (Rd)

1-1 3.00 2.24

1-2 3.16 2.31

2-1 5.86 3.27

2-2 5.93 3.30

The series 1 configuration exhibited an average ductility factor of 2.3; while for series 2, this

value is 3.3. Even though both system exhibit ductility, it is noticeable that series 2 is almost 1.4

times as ductile as series 1. So, in terms of desirable ductile failure mode for FFTT system; the

partially embedded connection with full embedment length is better than the fully embedded

connection with reduced embedment. The greater embedment length and larger lever arm for

series 2 caused less force transfer to the CLT and therefore, beam yielding occurred before any

significant damage to CLT. And in that series, after yielding, the system continued to pick up

load as steel beam undergo large post yield deformation.

The ductility factors obtained are not exact; rather, these are minimum ductility values for the

connection because the monotonic tests were discontinued upon reaching the peak load without

capturing the full degradation. Therefore, the deformation values used to calculate the connection

87

ductility are not the ultimate displacements. The ultimate displacements could be significantly

higher than the displacements at peak load. So, the actual ductility values for the connections are

expected to be higher than the values as listed in Table 14.

4.6.4 Hysteretic Behavior

The results of the cyclic tests are summarized in Table 15. It is noticeable that during the reverse

cyclic tests, series 1 was subjected to a greater peak moment (43.8 kN-m) than series 2 (38.5 kN-

m). Series 1 test was stopped after 45 load cycles while series 2 was discontinued after 53 cycles

due to beam uplifting from its longitudinal axis.

Table 15: Cyclic tests results

Series

Cyclic Tests

Peak Load

(kN)

Corresponding Peak Moment

(kN-m)

Number of

Cycles

Energy Dissipated (Joules)

LVDT-1 LVDT-2 LVDT-3 LVDT-4

1 62.3 43.6 45 75 187 N/A 2204

2 56.1 38.9 53 120 170 3230 6120

Series 1: The hysteretic curve obtained from LVDT-1 of series 1 test showed an initial flat

portion followed by sharp increase is load (Figure 62). This is due to a slight gap between beam

and CLT that was initially there due to fabrication error resulting in deformation without increase

in load. The slight arbitrary portion in otherwise a standard hysteretic curve obtained from

LVDT-2 can be attributed to erroneous reading (Figure 63). LVDT-3 readings were totally

erroneous and hence not considered. LVDT-4 produced a very well defined hysteretic curve

(Figure 64). These curves with their backbone are reproduced in Figures 75, 76 and 77.

88

Figure 75: Cyclic test: Series-1, LVDT-1 (with backbone curve)


‐60

‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐4 ‐3 ‐2 ‐1 0 1

Moment (kN‐m

)

Deformation (mm)

LVDT‐1

‐60

‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐5 ‐4 ‐3 ‐2 ‐1 0 1 2

Moment (kN‐m

)

Deformation (mm)

LVDT‐2

89


Series 2: The hysteretic curves obtained from LVDT-1 and LVDT-2 of series 2 which are inside

CLT showed well behaved hysteretic curves (Figures 70 and 71). However, there occurred some

sudden spikes in horizontal direction in these two curves. These happened due to lifting up of

beam from its longitudinal axis caused by out of plane buckling. LVDT- 3 and LVDT-4

produced well behaved hysteretic curves. These curves with their backbone are reproduced in

Figures 78, 79, 80 and 81.

‐60

‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐60 ‐50 ‐40 ‐30 ‐20 ‐10 0 10 20 30 40

Moment (kN‐m

)

Deformation (mm)

LVDT‐4

90



‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐12 ‐10 ‐8 ‐6 ‐4 ‐2 0 2 4 6

Moment (kN‐m

)

Deformation (mm)

LVDT‐1

‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐10 ‐8 ‐6 ‐4 ‐2 0 2 4

Moment (kN‐m

)

Deformation (mm)

LVDT‐2

91



The backbone curves obtained can be used to develop nonlinear hinge properties of the

connection for dynamic analysis of structure at global level. These can also be used to study

seismic performance criteria and checking suitability of such system.

‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐40 ‐30 ‐20 ‐10 0 10 20 30 40

Moment (kN‐m

)

Deformation (mm)

LVDT‐3

‐50

‐40

‐30

‐20

‐10

0

10

20

30

40

50

‐80 ‐60 ‐40 ‐20 0 20 40 60

Moment (kN‐m

)

Deformation (mm)

LVDT‐4

92

4.6.5 Energy Dissipation

For both series, energy dissipation occurred mainly through yielding of the steel beams. Very

little energy dissipation occurred inside the CLT. This is due to the fact that very little

deformation occurred inside the CLT during cyclic test while the beam underwent large post

yield deformation. The energy dissipated through different locations are reported in Table 15.

During testing of series 1, LVDT-1 and LVDT-2 dissipated 74.7 and 187.3 Joules of energy

while through beam yielding, 2204 Joules of energy were dissipated. The readings from LVDT-3

were erroneous during testing of series 1 and therefore not considered. Series 2 dissipated

significantly higher energy than series 1 at all LVDT locations. Maximum energy dissipated

during testing of this series was 6120 Joules which is almost 2.8 times the value obtained from

series 1. This is due to the fact that, large deformation was observed with ductile behavior in case

of series 2. So, series 2 performed better than series 1 and should be considered for further

studies. The amount of energy dissipation for both series is shown in Figure 82.

Figure 82: Energy dissipation during reverse cyclic tests

0

1000

2000

3000

4000

5000

6000

1 2 3 4

Energy Dissipated (Joules)

LVDTs

Series1

Series2

93

Chapter 5: CONCLUSIONS

5.1 Summary

This research focused on the component level performance of the steel beam to CLT panel

connection of the proposed hybrid FFTT system under quasi-static monotonic and reversed

cyclic loads. The combined numerical and experimental work yielded following main results:

1) The numerical investigation included the modelling of five previously tested configurations

and simulated the load-displacement behavior obtained by Bhat (2013). The numerical and the

experimental curves were in good agreement with the numerical curves being slightly stiffer.

This is due to the fact that in numerical modelling, fabrication imperfections were not considered

while these existed in the tested specimens. Nevertheless, the numerical model was deemed

appropriate to model the global deformation behaviour. On the material level, however, it was

shown that the linear-elastic timber model was insufficient to model the local plastic deformation

incurred in the timber, and, as a result, the obtained stress values, were unrealistic.

2) A numerical parameter study was conducted to recommend an improved connection geometry

which included steel bearing and side plates. Parameters of interest were embedment length and

depth, beam size and spacing of bearing plates. It was found that even the smallest wide flange

beam could cause excessive stresses and crushing in the CLT panel before yielding the beam.

3) The stress values from the numerical study indicate that the CLT panels were stressed beyond

yield and therefore, the linear elastic model was no longer sufficient. A nonlinear CLT material

model would simulate the post yield behavior more realistically. Such models can be developed

as mentioned in Chapter 2 (Grosse and Rautenstrauch, 2004). But post yield stress-strain data is

required for CLT which is currently not available.

94

4) Experiments (two quasi-static monotonic and one reversed cyclic test per series) were

conducted on two improved connection layouts. The first series consisted of fully embedded

(102 mm deep) wide flange beam with 350 mm spacing between bearing plates. The second

series consisted of partially embedded (85 mm deep) wide flange beam with 665 mm spacing

between bearing plates. The monotonic tests resulted in little damage and cracking to the panel

before steel yielding. The cyclic test on series 1 led to rolling shear failure in the weak CLT

layer. This failure mode was avoided in series 2 by partially embedding the beam and increasing

the spacing between bearing plates.

5) The experimental studies showed that both the system exhibit reasonable ductility with series

2 exhibiting higher ductility. The ductility factors were 2.28 and 3.29. These ductility factors are

minimum values established based on deformations at peak load rather than ultimate

deformations. The actual ductility for both systems are expected to be significantly higher.

6) The cyclic tests revealed that for both systems energy dissipation occurred mainly through

yielding of beam with very little dissipation happening inside CLT. This is expected and

desirable. The much higher energy dissipation of series 2 makes that series 2 better suited.

7) The load-deformation curves obtained from the study can be used to develop backbone curves

of the connections to define plastic hinge properties for nonlinear modeling of the FFTT system.

8) Overall, the study concludes that by using a partially embedded connection configuration with

large distance between bearing plates, the connection performance can be improved. However,

choosing the smallest steel section is not practical from a constructional point of view, where

significantly bigger sections would be required. If using bigger sections do not result in the

desired performance; then CLT may not be the ideal material for the FFTT system and LVL may

become a better option.

95

5.2 Recommendation for Further Studies

Future studies that can advance the knowledge on timber-steel hybrid systems include:

Developing nonlinear stress-strain curve for CLT panel for numerical modelling as elastic

strength properties do not account for post yield inelastic behavior.

A finite element numerical model with nonlinear timber properties which might simulate

better behaviour of the current system.

Conducting experiments with larger wide flange beams with different trial configurations by

varying embedment length and depth, CLT layer thickness, bearing area and distance etc.

Numerical and experimental investigation can be conducted with other mass timber products

which are stronger than CLT like LVL.

A wall testing program that includes static pushover as well as time history analysis to

observe system level behavior of the FFTT system.

96

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