New technologies of steel bridges in Japan

8/13/2019 New technologies of steel bridges in Japan

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Journal of Constructional Steel Research 58 (2002) 2170www.elsevier.com/locate/jcsr

New technologies of steel bridges in JapanT. Kitada a,* , T. Yamaguchi a, M. Matsumura a , J. Okada b ,

K. Ono c, N. Ochi daOsaka City University, Bridge Engineering Laboratory, Department of Civil Engineering, 3-3-138

Sugimoto, Sumiyoshi-ku, Osaka 558-8585, Japanb Department of Civil Engineering, Osaka City University, Osaka, Japan

c Department of Civil Engineering, Osaka University, Osaka, Japand Department of Civil Engineering, Akashi National College of Technology, Hyogo, Japan

Abstract

Introduced in this paper are (1) some computer programs for static/dynamic elastoplastic

nite displacement analyses used in designing steel bridge structures, (2) the ultimate strengthand design methods of steel plates, stiffened plates and columns made of high strength steelsubjected to compression, (3) seismic design and retrotting methods of new and existing steelbridge piers after the Hyogo-ken Nambu Earthquake, and (4) friction type joints and tensiontype joints with high strength bolts and development of high performance high strength boltsin Japan. These are some topics which the authors are concerned with among new technologiesof steel bridges under development in Japan, such as development of high performance steel,new ber materials, new types of bridges, performance based design methods, seismic designmethod against earthquakes of level 2 like the Hyogo-ken Nambu Earthquake, and bridgemanagement system. 2002 Elsevier Science Ltd. All rights reserved.

Keywords: Steel bridge; Elastoplastic nite displacement analysis; Dynamic analysis; High strength steel;Ultimate strength; Design method; Bridge pier; Seismic design; Retrotting method; High strength bolt;Friction bolt joint; Tension bolt joint

1. Introduction

Various kinds of new technologies are developed and investigated in the eld of steel bridges in Japan for reduction of construction cost and development of bridge

* Corresponding author. Tel.: +81-6-6605-2734; fax: +81-6-6605-2765. E-mail address: [email protected] (T. Kitada).

0143-974X/02/$ - see front matter 2002 Elsevier Science Ltd. All rights reserved.PII: S 0143-9 74X(01) 00029- 3


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22 T. Kitada et al. / Journal of Constructional Steel Research 58 (2002) 2170

engineering technologies. These are (1) development of high performance steel, suchas high strength steel, low yield ratio steel, extremely low yield stress steel, highelastic modulus steel, extremely thick steel plates, tapered steel plates longitudinallypro led, weathering steel, high ductility steel, etc., (2) utilization of new bermaterials, like glass ber, carbon ber and aramid ber, (3) development of newtypes of bridges, such as low cost bridges, new types of composite bridges, steel concrete mixed bridges, extremely long span bridges, etc., (4) investigation of per-formance based and limit state design methods, (5) development of rational andeconomical seismic design method of steel bridge structures against earthquakes of level 2 like the Hyogo-ken Nambu Earthquake, (6) introduction of fatigue designmethods into the design of steel bridges, and (7) development of rational bridgemanagement system.

Among these new technologies, rst of all, introduced in this paper are computerprograms, EPASS, USSP, EPASS Plus and USSP D for advanced static/dynamicelasto plastic nite displacement analyses used in designing steel bridge structuresidealized as rigid amed structures, plated structures and hybrid structures consistingof them.

Then, we describe the ultimate strength and design method of steel plates withall the edges simply supported, and steel plates with three edges simply supportedand the other unloaded edge free, steel plates with one longitudinal stiffener, andcolumns with box cross sections made of high strength steel subjected to com-pression. The method for deciding the correct safety factors of these members is

also discussed.The next topics are the seismic design and retro tting methods of new and existing

steel bridge piers after the Hyogo-ken Nambu Earthquake. They are key pointsobtained on the seismic design of steel bridge piers from the various kinds of damagedue to the Earthquake, the ductility design method after the Earthquake, and seismicretro tting methods to increase the ductility of existing steel bridge piers, keepingtheir ultimate strength almost constant.

The nal topics are the state-of-the-art friction type joints and tension type jointswith high strength bolts and development of high performance high strength boltswith high ductility in Japan.

In Japan, Japanese Speci cations for Highway Bridges (JSHB) are still describedby the format of allowable stress design method, although many long span bridges,for example, Akashi Kaikyou Bridge, the longest suspension bridge, Tatara Bridge,the longest cable-stayed bridge were constructed by using new and high technologiesof bridge engineering in Japan. However, general principals for the limit state designmethods of steel and concrete structures [1] have been already established, and thedesign codes for steel structures [2], composite structures [3] and the design standardsfor steel structures and steel concrete structures in railways have been publishedaccording to the design codes in the eld of civil engineering in Japan. For this

situation, this paper refers to not only the ultimate strength but also the allowablestrength and the safety factor to derive the allowable strength. The discussion on thesafety factor can be useful also for readers throughout the world.


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24 T. Kitada et al. / Journal of Constructional Steel Research 58 (2002) 21 70

Table 1Steel bridges of which load carrying capacity was checked by EPASS

Type of bridges Bridges Span length (m)

Framed bridge piers 1. Piers connected to spiral box 18.25 (height: 29.8 36.1)girder bridge, approach bridgefor Kizu River Br. (2 and 3stories)2. Bridge pier in Hanshin 18.0 (height: 36.2)Expressway (2 stories)

Cable-stayed bridges 3. Sugahara-Shirokita Br. (multi- 119 +238 +119cable type)4. Yamato River Br. (harp type 149 +355 +149cables)

Nielson Lohse bridges 5. Shin-Hamadera Br. (long 254span)6. Nakajima River Br. (wide 156.8 (clear width: 27.26)clear width)

Balanced arch bridges 7. Kizu River Br. (central part of 305Nilesen-Lohse type)8. Kishiwada Br. (vertical cables 255as hangers)

Suspension bridges 9. Konohana Br. (self-anchored 120 +300 +120mono-code cable)10. Tower of Kurushima Br. in height: 178Honshu-Shikoku Bridges

Curved box girder bridge 11. Approach spiral bridge for 42.0 +53.7 +1452.4 +53.3Kizu-River Br. (17 spanscontinuous box girder)

Continuous truss bridge 12. Ikitsuki Br. (long span) 200 +400 +200

5. Development of a seismic design method of steel bridges against strong earth-quakes like the Hyogo-ken Nambu Earthquake on the basis of the analysis byEPASS [18].

2.2. Elasto plastic and nite displacement analysis for plated structures

2.2.1. USSPUSSP is a program for the static, elasto plastic and nite displacement analysis

of thin-walled structural members idealized as plated structures with stiffeners. USSPis formulated on the basis of nite element methods. In USSP, plated structures areidealized as assembled models consisting to triangular elements and at beam col-umn elements (stiffener elements). USSP is used for investigating the ultimatestrength of thin-walled structural members under various combinations of loads and

enforced displacements for design or for research purpose. The details of the programare described in Refs. [19] [21].

USSP has the following main features:


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25T. Kitada et al. / Journal of Constructional Steel Research 58 (2002) 21 70

1. Practical residual stress distribution can be input automatically into analyticalmodels.

2. The behavior in the region of negative stiffness after the ultimate state of analyticalmodels can be simulated.

3. The veri cation of the program is suf ciently checked through the comparisonwith experimental results, elastic theoretical solutions and other numerical results.

The following researches and practical investigations were carried out by usingUSSP:

1. Development of the design method [22] for stiffened plates subjected to bi-axialin-plane forces on the basis of numerical results by using an analytical model

shown in Fig. 2.2. Investigation of the cause of buckling damage to stiffened plates with opening insteel bridge piers due to the Hyogo-ken Nambu Earthquake and repair methodsfor them [23].

3. Investigation of the ultimate strength of stiffened box girders under bending andtorsion [24], and stiffened box short columns subjected to compression [25], bi-axial bendings and torsion of which analytical model is illustrated in Fig. 3.

Fig. 2. Analytical model for stiffened plates subjected to bi-axial in-plane forces.


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Fig. 3. Analytical model for stiffened box short column.

4. Lateral torsion buckling and ultimate strength of horizontally curved I-girders[26].

5. Ultimate strength and design method of compression plates made of high strengthsteel [27].

2.2.2. EPASS Plus

EPASS Plus [28] [30] is a program for analyzing the static interactive behaviorof the overall buckling of bridge structures, the beam column buckling of structuralmembers and the local buckling of component stiffened plates through combiningUSSP with EPASS based on substructure method as shown in Fig. 4.

The following researches and practical investigations were carried out by usingEPASS Plus:

Fig. 4. Example of analytical model for EPASS Plus.


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Fig. 5. Rigid framed bridge pier damaged due to the Hyogo-ken Nambu Earthquake [31].

1. Analysis of the interactive behavior of the column buckling and the local bucklingof component plate panels of thin-walled box columns (see Figs. 5 and 6).

2. Analysis of the interactive behavior of the overall buckling and the local bucklingof component plate panels of the thin-walled box columns or the lower horizontalbox beam in a rigid framed bridge pier with two stories of which three web panelsin the central part of the lower horizontal box beam of this pier buckled in sheardue to Hyogo-ken Nambu Earthquake, as shown in Fig. 5.

Two models for analyzing the elasto plastic and nite displacement behavior of a long column with thin-walled box cross section by USSP and EPASS Plus areillustrated in Fig. 6. The numerical results by USSP and EPASS Plus are compared

Fig. 6. Numerical results by EPASS, EPASS Plus and USSP. (a) Model for EPASS Plus; (b) Modelfor USSP; (c) Numerical result.


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with each other, together with the result without considering the local buckling of the plate panels by EPASS. It can be seen from the gure that the numerical resultsby EPASS Plus and USSP coincide almost with each other.

The analytical models for the rigid framed bridge pier of which three web panelsaround the central part of the lower horizontal beam was damaged in shear bucklingdue to the Hyogo-ken Nambu Earthquake are shown in Fig. 5. One is the EPASSPlus model of which the central part of the lower horizontal beam is idealized withplate elements, and the other one is the EPASS Plus model of which the upper leftcorner part is idealized with plate elements. The numerical results by EPASS Plusare shown in Fig. 7(a) compared with each other, together with the result withoutconsidering the local buckling of the plate panels by EPASS. In this gure, a is theload parameter, multiplier to the design seismic load. It can be seen from Fig. 7(a)that the numerical results by EPASS Plus and EPASS are almost the same. Thedeformation of the USSP block of the model (beam) at the ultimate state is illustratedin Fig. 7(b). This deformation is similar to the actual failure mode.

2.3. Elasto plastic, nite displacement and dynamic analysis

A computer program USSP D [32], has been already developed for simulating theelasto plastic, nite displacement and dynamic response of steel bridge piers ideal-ized into a vibration system with single mass by considering the local buckling of the component stiffened plate panels through combining a computer program USSP

[33] for analyzing the elasto plastic and nite displacement behavior of stiffenedplates with a computer program FDDA1 (FDM) [34] for analyzing the dynamicresponse of vibration systems with single mass. The brief owchart of the programUSSP D is shown in Fig. 8.

As a numerical example, a cantilever steel column with box cross section isadopted in this section. For the purpose of the veri cation of USSP D the numerical

Fig. 7. Numerical results of rigid framed bridge pier by EPASS and EPASS Plus. (a) Numerical results;(b) Deformation of central part in lower horizontal box beam.


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Fig. 8. Flowchart of USSP D.

results through USSP D are compared with the experimental results by the pseudo-dynamic tests [35]. The applied seismic acceleration for the analytical model is 1.5times the acceleration measured at the Higashi Kobe Ohashi Bridge (the type IIIground, wear ground, maximum acceleration: 488.8gal) in the Hyogo-ken Nambu

Earthquake. It can be seen from Fig. 9 that the dynamic behavior obtained byUSSP D is similar to the experimental results by the pseudo dynamic tests [35].The deformations of the analytical model at representative times are illustrated in

Fig. 10. It can be observed from this gure that the parts of the columns in thevicinity of the bases of the models are remarkably buckled.

3. Design method of bridge members made of high strength steel

Mild steel has clear yield plateau, while high strength steel does not, and the

stress strain curves for these materials are very different from each other. And theshape of residual stress distribution in component plate panels of built-up membersmade of high strength steel is also different from that of mild steel as shown in Fig.


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Fig. 9. Time history of horizontal displacement. (a) Numerical result; (b) experimental result.

Fig. 10. Deformations at representative times (Displacement: 10). (a) t =5 sec; (b) t =6 sec; (c) t =7 sec;(d) t =8 sec; (e) t =9 sec; (f) t =29 sec.

11. In the Japanese current design method, however, the ultimate strength, s u , of compression members made of high strength steel and mild steel is evaluated fromsame ultimate strength curves derived as completely elasto plastic material of which

yield stress is, s Y . For high strength steel, 0.2% strength, s 0.2 Y , is used as the yieldstress. And the allowable compressive stress in the Japanese Speci cations for High-

Fig. 11. Residual stress distribution in plate panels. (a) High strength steel; (b) mild steel.


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Fig. 13. Division of nite elements and boundary conditions (simply supported plates).

Rbt s Y E 12(1 m

2 )p 2 k

(3)

where b is the width, t is the thickness, s Y is the yield stress, E is Young s modulus, m is Poisson s ratio and k is the buckling coef cient ( k =4) of the plates. The initialdeection mode of the simply supported plates is given by

w0 w 0 cos( p x / b)cos( p x / a ) (4)

where w 0 = b /150 according to JSHB [36]. In the analysis, by referring to Ref.[37],the residual stress distributions shown in Figs. 14 and 15 are considered for highstrength steel and mild steel, respectively.

An outstanding (projecting) plate with three edges simply supported and one freeedge subjected to uniform compressive displacement is the analytical model and isshown in Fig. 16. The boundary conditions are imposed as shown in Fig. 17. Andhalf of the plate is extracted and used in the analysis for the symmetry of the plateshown in Fig. 17. In this study, the following parametric analysis is executed. Theplate slenderness parameter R given by eq. (5) is changed in the region from 0.3 to2.0, (=a / h) from 1 to 4.

Rht s Y E 12(1 m

2 )p 2 k

(5)

where h is the width and k is the buckling coef cient ( k =0.425). The initial de ectionmode adopted in the analysis is given by the following equation.

w0 w 0 / h y cos(p x / a ) (6)

where w 0 / h = 1/100 according to JSHB [36]. In the analysis, by referring to Refs.


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Fig. 14. Residual stress distribution (simply supported plates of high strength steel (HT785 and HT685)).

Fig. 15. Residual stress distribution (simply supported plates of mild steel (SM400)).

[37] and [38], the residual stress distributions shown in Figs. 18 and 19 are con-sidered for high strength steel and mild steel, respectively.

A continuous stiffened plate with one longitudinal stiffener subjected to uniformcompressive displacement is used in the analysis, as shown in Fig. 20. And a partdivided by nite elements can be extracted and used in the analysis for the symmetry.The state of mesh division of nite elements is shown in Fig. 21 and Table 2. Thefollowing parametric analysis is carried out considering the residual stress and initialdeformation. The plate slenderness parameter R given by eq. (7) is varied from 0.25to 1.5, (=a / h) from 1 to 4.

Rb lt s Y E 12(1 m

2 )p 2 k

(7)


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Fig. 16. Analytical model (outstanding plates).

Fig. 17. Division of nite elements and boundary conditions (outstanding plates).

where bl is the spacing of longitudinal stiffener and k is the buckling coef cient(k =4). And in the stiffened plate, the exural rigidity ratio of longitudinal stiffenerto plate panel, g l, equals the required minimum value, g lreq , dened by JSHB [36].The width thickness ratio of the longitudinal stiffener, hr / t r , is 7 for preventing thelongitudinal stiffener from buckling locally. Besides, the transverse stiffeners arevery rigid in the analytical models and the effect of the transverse stiffeners is con-sidered by regarding the location of the transverse stiffeners as simple supports. The

initial de ection modes used in the analysis are shown in Table 2. In the analysis,according to Ref. [37], the residual stress distribution shown in Fig. 22 is assumed.And the values of the residual stress listed in Table 3 are used.


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Fig. 18. Residual stress distribution (outstanding plates of high strength steel (HT785 and HT685)).

Fig. 19. Residual stress distribution (outstanding plates of mild steel (SM400)).

Fig. 20. Analytical model (stiffened plates).

3.1.2. Stress strain curves used in analysis

The stress strain curve of high strength steel (HT785 and HT685) used in theanalysis is shown in Fig. 23. And the stress strain curve of mild steel (SM400) isalso shown in Fig. 23.


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Fig. 21. Division of nite elements and boundary conditions (stiffened plates).

Table 2Mesh division of nite elements and initial de ection modes used in the analysis a

Aspect ratio Mesh division of Initial de ection modesnite elements

1.0 18 12 w0 =d 0cos( p x/a)sin( p y/b)+w10 cos(3 p x/a)sin( p y/bl)2.0 30 12 w0 =d 0cos( p x/a)sin( p y/b)+w10 cos(5 p x/a)sin( p y/bl)3.0 42 12 w0 =d 0cos( p x/a)sin( p y/b)+w10 cos(7 p x/a)sin( p y/bl)4.0 54 12 w0 =d 0cos( p x/a)sin( p y/b)+w10 cos(9 p x/a)sin( p y/bl)

a d 0 =a /1000; w10 =b l /150 according to JSHB.

Fig. 22. Residual stress distribution (stiffened plates). (a) Plate panel; (b) stiffener.


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Table 3The values of residual stress a

Variation of s rt s rc s rc,r s r,r steel/residual stress

HT785 0.80 0.15 0.12 0.60HT685 0.80 0.15 0.12 0.60SM400 1.00 0.30 0.20 0.60

a The values non-dimensionalized by the yield stress, s g .

Fig. 23. Stress strain curves used.

3.1.3. Analytical resultsIt can be seen from Fig. 24 that the ultimate strength of simply supported plates

made of high strength steel is higher than that of mild steel by the maximum value

Fig. 24. Comparison of ultimate strength curve of high strength steel with that of mild steel (simplysupported plates).


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Fig. 25. Comparison of ultimate strength curve of high strength steel with that of mild steel (outstandingplates).

of about 10% at R nearly equal to 0.8, because the compressive residual stress non-dimensionalized by the yield stress of simply supported plates made of mild steelis larger than that of high strength steel. In the case where R is smaller than 0.4,the ultimate strength of simply supported plates made of high strength steel is higherthan the yield stress s Y , because of strain hardening of high strength steel as shownin Fig. 23. The strength curve of JSHB seems to be very different from the ultimatestrength curve through the numerical results.

According to Figs. 25 and 26, similar things to the analytical results of simplysupported plates can be seen in the outstanding plates and stiffened plates.

Fig. 26. Comparison of ultimate strength curve of high strength steel with that of mild steel (stiffenedplates).


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3.1.4. Design method of simply supported, outstanding and stiffened plates madeof high strength steel

As shown in eq. (1), it is de ned in JSHB [36] that the smaller value of s B /2.2and s u /1.7 is the allowable compressive stress of the plates made of high strengthsteel, while the allowable compressive stress of the plates made of mild steel iss u /1.7. Firstly, for investigating the validity of the design method mentioned above,the ultimate strength curve of outstanding plates made of high strength steel throughthe analysis, an allowable stress curve which is the ultimate strength curve dividedby the safety factor 1.7, and the tensile strength s B of high strength steel non-dimen-sionalized by the yield stress s Y and then divided by the safety factor 2.2 are shownand compared in Fig. 27. In the calculations for depicting this gure, the values of 785 N/mm 2 and 685 N/mm 2 dened in the design speci cations for bridge superstruc-tures of Honshu-Shikoku Bridge Authority [39] are used as the values of s B and s Y of high strength steel HT785, respectively. It can be seen from this gure that theallowable stress is almost decided by s B /2.2 in the region of R 1.0 according tothe design concept of JSHB and by the ultimate strength curve generated throughthe analysis divided by 1.7 in the region of R 1.0.

Furthermore, the relationships between the average compressive stress s and thedeection of the central point on the free edges of two outstanding plates made of high strength steel HT785 are compared in cases of R=0.7 and 1.0 in Fig. 28. Noevident difference can be observed in the behavior of these two compression plates.However, according to the design method of JSHB, as shown in Fig. 27, the safety

factors for the compression plates with R=

0.7 and 1.0 are 1.9 and 1.7, respectively,and these values are different. No reason can be found out from this gure in chang-ing the safety factors of these two outstanding plates. For the reason mentionedabove, it is seemed to be rational that the safety factor of compressive plates madeof high strength steel is 1.7 in all the regions of the plate slenderness parameter R.The allowable compressive stress of plates made of high strength steel can be, there-fore, proposed by the following equation.

s ca s u /1.7 (for high strength steel and mild steel) (8)

Fig. 27. Ultimate strength and allowable compressive stress curves (outstanding plates made of highstrength steel HT785).


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Fig. 28. Relationship between average stress and de ection of outstanding plates made of high strengthsteel HT785 under compression ( a =1).

The following design strength curves are derived as the regression curves of theultimate strength curves of simply supported plates through the analysis.

Design strength curve proposed in this study (high strength steel):

s u / s y

1.0 ( R 0.5

0.58 R 1.29 (0.5 R 1.0

0.7/ R0.75 (1.0 R 2.0

(9ac)

Design strength curve proposed in this study (mild steel):

s u / s y

1.0 ( R 0.35)

0.520 R 1.182 (0.35 R 1.0)

0.662/ R0.65 (1.0 R 2.0)

(10a c)

These design strength curves are compared in Fig. 29.

Fig. 29. Strength curves of simply supported plates made of high strength steel.


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The following design strength curves are derived as the regression curves of theultimate strength curves of outstanding plates through the analysis.


s u / s y

1.0 ( R 0.7)

0.5 R 1.35 (0.7 R 1.2)

0.807/R 0.4 (1.2 R 2.0)

(11a c)


s u / s y

1.0 ( R 0.55)

0.480 R 1.264 (0.55 R 1.0)

0.78/R 0.45 (1.0 R 2.0)

(12a c)

These design strength curves are compared in Fig. 30. The following designstrength curves are derived as the regression curves of the ultimate strength curvesof stiffened plates through the analysis.


s u / s y

1.0 ( R 0.45)

0.640 R 1.288 (0.45 R 1.0)

0.65/R0.9

(1.0 R 1.5)

(13a c)


s u / s y

1.0 ( R 0.3)

0.40 R 1.12 (0.3 R 0.5) 0.66/R 1.25 (0.5 R 1.0)

0.59/ R0.75 (1.0 R 1.5)

(14a d)

Fig. 30. Strength curves of outstanding plates made of high strength steel.


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Fig. 31. Strength curves of stiffened plates made of high strength steel under compression.

Fig. 32. Analytical model for columns with unstiffened box cross section.


3.2. Column members

3.2.1. Analytical modelsThe column with unstiffened box cross section subjected to compressive force is

the analytical model as shown in Fig. 32. The column has both the ends simply

supported, and half of the column is analyzed for the symmetry. The plate slender-ness parameter R given by eq. (3) of the column is 0.5, so that the local bucklingof the component plate panels does not occur. The thickness of the plate panels iscalculated from R=0.5 and listed in Table 4. The following parametric analysis con-

Table 4Dimensions of cross section

Types of steel/Dimensions of Width b (cm) Thickness t (cm)cross section

HT785 28.0 1.81HT685 28.0 1.69SM400 28.0 1.00


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Fig. 33. Analytical model for columns with stiffened box cross section.

sidering the residual stress and initial deformation is carried out. The slendernessparameter l given by the following eq. (15) is changed in the region from 0.05 to 2.0.

l1p s Y E ler (15a,b)

where le is the effective buckling length and r is the radius of gyration. The initialdeection mode of the column model is given by

W 0 W 0 cos( p Z / l) (16)

where W 0 = l/1000 according to JSHB. By referring to Ref. [37], the residual stressdistributions shown in Fig. 22 and the values of residual stress listed in Table 3are considered.

A column with stiffened box cross section and with both the ends simply supportedis the analytical model as shown in Fig. 33. Half of the column model can be usedin the analysis for the symmetry. The plate slenderness parameter R given by eq.(7) of the column model with stiffened box section is 0.5 for the same reason asthe analysis of the column model with unstiffened box section. The thickness of theplate panels is calculated from R=0.5 and is shown in Table 5. The exural rigidityratio of the longitudinal stiffeners to the plate panels, g l, equals the minimum required

value, g lreq , dined by JSHB [36]. The width thickness ratio of the stiffeners, hr / t r ,Table 5Dimensions of cross section

Dimensions of cross section

Plate panel Stiffener

Type of steel Width b (cm) Thickness t (cm) Width h (cm) Thickness t (cm)

HT785 28.0 0.91 7.20 1.03HT685 28.0 0.84 6.74 0.96SM400 28.0 0.50 4.24 0.61


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is 7 for preventing them from local buckling. The following parametric analysisconsidering the residual stress and initial deformation is executed. The slendernessparameter l is changed in the region from 0.05 to 2.0. The initial de ection modeis given by eq. (16). The residual stress distributions shown in Fig. 22 and the valuesof residual stress in Table 3 are considered.

In the analysis of the column models, the stress-strain curves shown in Fig. 23are used.

3.2.2. Analytical resultsIt can be seen from Fig. 34 that the ultimate strength curves of columns with

unstiffened box section made of high strength steel (HT785 and HT685) are higherthan those of mild steel (SM400) at R nearly equal to 1.0, because the compressiveresidual stress non-dimensionalized by the yield stress s Y of columns made of mildsteel is larger than that of high strength steel as shown in Table 3. In the regionwhere l is smaller than 0.1, the ultimate strength of the columns with unstiffenedbox section made of high strength steel is higher than the yield stress s Y due to thestrain hardening of high strength steel as shown in Fig. 23. The strength curve of JSHB is similar to the ultimate strength curve of columns with unstiffened box sec-tion made of mild steel through the analytical results.

According to the analytical results shown in Fig. 35, similar things to the analyticalresults of columns with unstiffened box section can be also seen. And it can beobserved from Fig. 36 that the ultimate strength curves of the columns with stiffened

box section made of high strength steel and mild steel are similar to those of columnswith unstiffened box section made of high strength steel and mild steel, respectively.

3.2.3. Design method of columns with unstiffened and stiffened box section madeof high strength steel

The safety factor and allowable stress of columns made of high strength steel aredened as eq. (1). Fig. 37 is depicted through a similar method shown in Fig. 27.

Fig. 34. Ultimate strength curves of high strength steel and mild steel (columns with unstiffened boxsection).


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Fig. 35. Ultimate strength curves of high strength steel and mild steel (columns with stiffened boxsection).

Fig. 36. Ultimate strength curves of columns with unstiffened box section and stiffened box section.

It can be recognized from Fig. 37 that the safety factor in the region of l 0.5 isdifferent from that of l 0.5. However, it is considered to be rational that thesafety factor of columns made of high strength steel is 1.7 in all the region of theslenderness parameter l for the same reason mentioned in 3.1(4). The allowablecompressive stress of columns made of high strength steel can be therefore proposedby eq. (8).

The following design strength curves of columns with box cross section arederived by the least squares method to t the ultimate strength curves through theanalysis.


s u / s y1.0 ( l 0.25)0.006 l 2 0.718 l 1.118 (0.25 l )

(17a,b)


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Fig. 37. Ultimate strength and allowable compressive stress curves (columns with stiffened box sectionmade of high strength steel).


s u / s y1.0 ( l 0.25)0.118 l 2 0.718 l 1.172 (0.25 l )

(18a,b)


4. Seismic design and retrotting of steel bridge piers

4.1. Seismic design of steel bridge piers

Steel structures, compared with concrete ones, are light and strong. Moreover,steel structures can be built even under severe constructional conditions, for instance,

Fig. 38. Strength curves of columns with box cross section made of high strength steel.


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in narrow and limited spaces at urban areas like Tokyo, Osaka, Nagoya, Fukuokaetc in Japan. As various erection methods can be considered and adopted in con-structing steel structures, steel bridge piers are, therefore, mainly constructed at sev-ere constructional locations where many spatial limitations including building regu-lations and the effective use of the limited spaces are desired strictly. Steel bridgepiers are also applied to locations where heavy supper structures are not unfavorable,for example, on soft ground, reclaimed land and bay areas.

4.1.1. Necessity for retro ttingSteel as one of structural materials is so ductile in comparison with concrete that

had been considered that the steel structures designed against earthquakes (level 1)with the maximum acceleration of 150 200 gals at the surface level of the ground,

never collapse against even strong earthquakes (level 2) which rarely occur duringtheir design life, although they may lose some of their functions. However, the 1995Hyogo-ken Nambu Earthquake (one of the level 2 earthquakes) caused destructivedamage to highway bridges. Steel bridge piers suffered heavy damage for the rsttime in Japan. The damage includes the collapse of two steel bridge piers with rec-tangular cross section. The earthquake provided a large impact on the earthquakedisaster preventions in various elds in Japan and revealed that there were a lot of critical issues to be revised in the seismic design methods of bridge structures. Afterthe earthquake, many energetic investigations for making steel bridge piers so ductilehave been carried out that they can support superstructures without increasing theirelastic strength against such a strong earthquake although their functions may belost [40]. On the basis of these research results, the Part V. Seismic Design, theJapanese Speci cations for Highway Bridges (JSHB) was totally revised in 1996[41], and the design procedure moved from the traditional seismic coef cient methodto the ductility design method. The following paragraphs show the outline of thebasic concept of the seismic design method of steel bridge piers provided in the1996 Seismic Design, Japanese Speci cations for Highway Bridges [42].

On the basis of the detail survey of damage to steel bridge piers [43] and thendings obtained from analyses and experiments on seismic design of steel bridgepiers [44], the following key points are regarded as the basic concepts of the seismicdesign of steel bridge piers.

4.1.2. Avoidance of speci c brittle failure modeOn the basis of the failure modes and deformation of steel bridge piers damaged

due to the Hyogo-ken Nambu Earthquake, the failure modes illustrated in Fig. 39should be avoided.

The illustration shown in Fig. 39(a) expresses one of brittle failure modes of steelbridge pier with rectangular section. Such a failure mode was observed in two steelbridge piers which completely collapsed in the Hyogo-ken Nambu Earthquake. In

this failure mode, vertical cracks along the welded corners of stiffened steel platepanels occurred after the occurrence of serious local buckling of stiffened steel platepanels. This would mean loss of the vertical strength against the dead weight of the


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Fig. 39. Typical brittle failure modes of steel bridge piers. (a) Steel bridge pier with rectangular section;(b) Steel bridge pier with circular section.

superstructure. If this failure mode happens, it can cause the sudden decrease in thebearing capacity against the dead weight of superstructures.

Fig. 39(b) shows a brittle failure mode of steel bridge piers with circular section.In this failure mode, concentrated and large deformation appears after the local buck-ling at a limited location, resulting in cracking around the column s circumferencedue to the increased deformation.

4.1.3. Improvement of ductilityIn the 1990 seismic design speci cations, the ductility design method to check that bridge piers do not collapse against the earthquake of Level 2 was applied toonly reinforced concrete bridge piers. However, it was decided to apply the ductilitydesign method also to steel bridge piers since damage caused by the Hyogo-kenNambu Earthquake suggested the importance of ensuring suf cient ductility of thestructural systems.

4.1.4. Restriction of residual displacement If the seismic design allows the excessive ductility in plastic region of steel bridge

piers, large displacement will occur accompanied with large residual displacementin bridge piers. Large residual displacement generated in the bridge piers will makethe restoration work dif cult to perform. For this reason, the seismic design basedon the ductility design method demands to restrict the residual displacement to repair-able limit.

4.1.5. Seismic design of anchorsThe anchor part of a bridge pier is an important structure that transfers the axial

force, bending moment and shear force from the steel bridge pier to the foundation,and has signi cant in uence on the seismic performance of the steel bridge pier. In

the view of the dif culty of identifying the damage to the anchor part and carryingout the large repair, the seismic design of anchors have been established in such away that the risk of the plastic deformation in the anchor part is minimized.


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4.2. Seismic retro tting of existing steel bridge piers

The 1996 seismic design method recommends that steel bridge piers should beencased with concrete as one of the most effective and economical means to preventtheir outer steel plates from local buckling. No recommendable structures seem tobe speci ed except for a rectangular cross section with the corners stiffened by ribplates with regard to ductile steel cross sections. On the other hand, it is requestedthat almost all the existing steel bridge pier column should be retro tted to achievethe required new design criteria.

Speci cations concerning a seismic retro tting design for the existing steel bridgepiers were drafted by Japan Road Association, Hanshin Expressway Public Corpor-ation [45] and the other organizations after the Hyogo-ken Nambu Earthquake. Inthese speci cations, two basic seismic retro tting methods are prepared to enhancethe ductility of existing steel bridge pier columns in case their basement structuresare not damaged:

1. Filling concrete into the column member of a bridge pier.2. Retro tting the component stiffened plate panels of the pier column by using

additional steel members, such as transverse stiffeners, longitudinal stiffeners,ange plates etc.

It is important to avoid the substantial increment of the ultimate strength of an

existing steel bridge pier column in retro tting the pier column. The former method(1) is not adaptable if the ultimate strength of the pier column lled concrete exceedsmore than that of its basement structure. In that case, the other method (2) is adoptedalthough this method is expensive and inconvenient in repair work.

4.2.1. Filling concrete into the column member of a bridge pier As con ned concrete in a hollow steel column can prevent the stiffened plate

panels from buckling inside the cross section, it is easily expected that the ductilityof the column is enhanced. Then this seismic retro tting method has a great advan-tage for practical and economical reasons. Its adoption to existing steel bridge pier

columns, therefore, is examined prior to the other retro tting methods.However, the substantial increase of the ultimate strength of the bridge pier columndue to the effect of the concrete lled inside of the cross section is apt to causeserious damage to the basement structure. As the replacement or retro tting of thebasement structure requires signi cant cost and time for repair works, and the ndingof the damage there is very dif cult, so this method cannot be a good retro ttingmethod. Accordingly more large-scaled basement structures are required not onlyfor newly constructed bridge piers but for retro tting existing bridge piers.

4.2.2. Retro tting component stiffened plate panels by steel members

To obtain expected ductility for an existing pier column without lling concreteinto it, the restrictive conditions concerning the plate slenderness parameters on thelocal buckling of its cross-section are regulated, for example in Ref. [46] as follows:


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Fig. 40. Example of seismic retro tting method in Ref. [45] (actual case).

The slenderness parameter of plate panels between longitudinal stiffeners, R R( = s y / s cr , s cr : elastic buckling stress of plate panels) is less than 0.4. The slenderness parameter of overall stiffened plate panels, RF ( = s y / s crg , s crg :elastic buckling stress of overall stiffened plate panels) is less than 0.4.

The slenderness parameter on the local buckling of longitudinal stiffeners, RS isless than 0.5.

The following stiffening methods are concretely adopted for the stiffened platepanels and the existing longitudinal stiffeners (see Figs. 40 and 41) to satisfy therestrictive conditions mentioned above:

The plate panel between the existing longitudinal stiffeners is stiffened with theadditional and small longitudinal stiffeners to ensure the required slendernessparameter for the plate panels.

The existing longitudinal stiffeners are still more stiffened with the additionalange plates to reduce the slenderness parameter of the stiffened plate panels andlongitudinal stiffeners themselves.

Fig. 41. Example of seismic retro tting method for stiffened plate panels.


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Moreover, the narrow gaps are introduced between both the ends of retro ttingmembers, the additional longitudinal stiffeners and the additional ange plates, andthe transverse stiffeners or diaphragms [47] to mitigate the strength increment of thepier column itself due to retro tting.

It is reported that this retro tting method can bring suf cient ductility to the exist-ing pier columns with 20% increase of the ultimate strength of the original piercolumns in Ref. [8]. That is because the existing longitudinal stiffeners and platepanels at the narrow gaps deform plastically prior to the other cross sections. How-ever, the method has problems to be solved from a practical and economic a pointof view, i.e., a lot of fabricating work inside of the box cross section, like welding,connecting, painting and so on, are needed in practices.

4.2.3. Concept for improved retro tting method for existing steel bridge piersIn this paragraph a new retro tting strategy for existing steel bridge piers is

developed to satisfy the seismic design criteria and to improve the current retro ttingmethods mentioned above is described. The following issues should be consideredfor the improved retro tting method of existing steel bridge piers:

1. The method is to be easy, practical and economical.2. Detection of damage is easy.3. Damage can be easily repaired.

Fig. 42 shows the outline of the improved retro tting method. The method requiresthat a short segment with the steel cross section in the column (hereafter calledductile segment) deform plastically prior to the other part of the column in order toenhance the ductility and to mitigate the increment of the ultimate strength due tothe retro tting. That is, the length of the ductile segment is shortened in order toprevent the pier column from the local buckling up to largely deformed region, to

Fig. 42. Outline of concrete- lled steel bridge pier column with short ductile segment.


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have enough ultimate strength but less than that of anchor bolts, and to increase theplastic deformation capacity only in the ductile segment.

By setting the ductile segment and designing it to have enough strength and duc-tility practically, the steel bridge pier itself can be considered to be able to supportits superstructure without any damage against earthquakes of Level 1, and not col-lapse due to earthquakes of Level 2 although the superstructure supported eventhough it cannot be used for traf c. The ultimate strength and ductility of the ductilesegment can be exactly estimated in the design, because the mechanical propertiesof steel materials, such as yielding stress, tensile strength and Yong s modulus arestable compared with those of concrete. That means the design methods, design toolsand experiences used before the Hyogo-ken Nambu Earthquake can be utilized inthe design of the other part of the column except for the ductile segment. Moreover,reconstruction of large-scaled basement structures or reinforcement of them is notneeded.

The ductile segment should never be placed in the lower part of the pier columnlocated underneath the ground but in the middle part of the pier column over theground. This makes it possible to nd damage easily and to estimate the extent of the damage to the pier column easily through rough inspections of change of paintingstate, local buckling and so on, of the damaged ductile segment. Moreover the dam-aged steel plates at the ductile segment is repairable into the original shape bypressing or replacing it to the new ones by a similar method used for restoring steelpier columns damaged due to the Hyogo-ken Nambu Earthquake. It should be noted

that composite cross section consisting of stiffened plates and encased concrete insideof the steel cross section has dif culties in repairing or replacing of the damaged part,because of inconvenient gas cutting caused by the existence of the encased concrete.

Fig. 43 illustrates various types of structural details for the ductile segment, whichare contrived by considering their practicality and workability.

These can easily generate the steel ductile segment, which deforms plasticallyprior to the other cross sections in the column. Only the proper length of the ductilesegment should be decided in such a way that the local buckling occurs at the ductilesegment at extremely lower load level.

Note that too short a length of the ductile segment results in the ultimate strength

of the pier column becoming larger than that of the anchor bolts because of strainhardening phenomenon of the ductile segment. Furthermore, the ductile segment withadequate length must be placed in such a way that the ratio, M a / M su of the appliedbending moment of, M a to the ultimate bending moment, M su of the ductile segmentis less than the ratio, M a * / M anku of the applied bending moment of the cross sectionof the anchor bolts, M a

* divided by the ultimate bending moment M anku , of the section(see Fig. 44).

5. Joints with high strength bolts in Japan

Two types of eld joints for structural members of steel bridges are used in Japan;one is welding type and the other is bolted type. In recent years, the welding type


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Fig. 43. Various types of structural details for ductile segment.

Fig. 44. Design concept of bridge pier columns with ductile segment.

joint is often used because of keeping the surface of connected members at froman aesthetic viewpoint. However, considering the easiness of construction in general,the bolted type seems to be superior to the welding type. The bolted type connectionis generally used for bridge structures as a typical method of connecting structuralmembers on site.

Japanese Speci cations for Highway Bridges (JSHB) [36] classify bolted connec-tions into three types, such as friction, bearing and tensile types. In JSHB, the frictiontype is mainly speci ed in detail as well as in other speci cations for steel structures

in the world. Although these design concepts are almost the same all over the world,Japanese speci cations may be a little more conservative than other speci cationsin the world.


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5.1. Friction type joints

5.1.1. Reconsideration of JSHB speci cations on friction type jointsThe JSHB speci cations on friction type joints are based on traditional speci -

cations of riveted joints. The JSHB speci cations on friction type joints are conserva-tive, because the actual mechanical behavior on friction type joints is different fromthat of riveted joints. In order to make the JSHB speci cations more rational, theinherent mechanical behavior of the friction type joints must be considered in theirdesign like other speci cations in the world, such as AASHTO, Eurocode and soon. For example, following issues should be solved:

1. Reconsidering the slip coef cient and effective tensile cross sectional area of bolts.2. Establishment of a rational method for evaluating the resistance bending moment

of connected girders subjected to bending through considering the interactionbetween the ange and web sections.

Nishimura and Akiyama propose a more rational design method paying attentionto the mechanical behavior of the friction type joints through the experimental andanalytical researches [48,49].

The slip coef cient, m proposed by them is shown in Fig. 45 and eq. (19). Thevertical axis shows the slip coef cient, and the horizontal axis shows the ratio, b of the nominal slip strength, N * SL to the nominal yield strength, N * Yn of the connected

members. It can be found that the slip coef cient decreases as the index value b islarger than 0.7, and the index value b in case that b keeps constant value, 0.5 issmaller than 0.7. The reason why the slip coef cient changes with the index value b is that the decrease of the bolt axial force due to yielding of the connected membersis considered.

Fig. 45. Slip coef cient by Nishimura and Akiyama.


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m 0.5 ( b 0.7) m 0.5 (1.28 0.4 b) (0.7 b 1.0)

(19a,b)

In addition, they also propose an evaluation method of the ultimate strength of thefriction type joints of plate girders. In JSHB speci cations, the safety of the frictiontype joints of plate girders subjected to bending is checked by the veri cation of each part, such as the ange section and the web section. That is, the overall ultimatestrength of the joint section including the ange sections and the web section is notevaluated in the JSHB speci cations. However, in fact, slip phenomena of a jointof a plate girder occurs as soon as both the web and the ange sections reach theirslip state, and it does not occur when either the web section or the ange sectionreaches the slip state. They also report that joints designed by the proposed procedurecould be more rational and economical than the JSHB design [50].

5.1.2. Recent research activities on friction joints in JapanIn Japan, some recent steel bridges seem to consist of thick plates without stiff-

eners due to reduction of fabrication cost. For example, thickness or width of angeplates of these plate girder bridges does not change throughout one member. As aresult, a connected section is not always rational from a viewpoint of economicalaspect, because the cross sectional forces are not high compared with the ultimatestrength of the section. Longitudinal pro led plates, namely longitudinally taperedplates, may be used for the lower and upper ange plates in order to improve thissituation. And the mechanical behavior and rational design method for such frictiontype joints as illustrated in Fig. 46 are investigated experimentally and analyticallyby Nishimura and Kamei for practical use of such joints. They conclude that such joints can be applicable without pre-bending of the connection plates as shown in Fig.46 and that the design slip coef cients need not to be changed for such joints [51].

Furthermore, in the case of joints for plate girders with much thicker ange plates,it is dif cult to join the members only by high strength bolts. Because there are nosuch longer high strength bolts in ordinal. In such case, the ange and web sectionsmight be connected by welding and high strength bolts respectively. In these typeof joints, the fabrication procedure is very and shrinkage of welding must be con-

sidered [52].

Fig. 46. Typical examples of friction type joints of longitudinal pro led plates.


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5.2. Tension type joints

5.2.1. Classi cation of high strength bolted tensile jointsHigh strength bolted tensile joints carry cross sectional forces occurring at the

interface between connected members by reduction of contact stress due to the pre-stressed tensile force in the bolts. This type of tensile joints can transfer larger loadsper bolt than the friction type. This means that the number of the bolts required ata connection can be reduced, and consequently ease of fabrication and constructioncan be expected. In addition, because of introducing high tensile force into the bolts,this type has some other advantages, such as high rigidity and high fatigue strength.In case of a tensile joint subjected to shearing force, the joint can act as the frictiontype joint if no gap occurs between the connected sections.

Generally speaking, tensile joints are classi ed into two types according to thelength of bolts. One type is called the short connection type with short bolts asshown in Fig. 47(a). The short connection type is mainly used for connecting beamsand columns in steel building structures. The other type is called the long connectiontype consisting of long bolts and additional stiffeners. A typical example of this typeof joints is illustrated in Fig. 47(b). In the long connection type of joints, additionalbolt axial force due to prying action can be mitigated by using long bolts. Featuresof each type of joints are summarized in Table 6. The evident difference betweenthem is the existence of the prying action. If the additional bolt axial force due tothe prying action can be estimated precisely, the short connection type is superior

to the long connection type for ease of fabrication and construction, although theshort connection type of joints may not be superior to the long one in ductility.

5.2.2. Examples of tensile joints in JapanIn Japan, tensile joints had not been used for not only primary members but also

secondary members until the JSSC recommendation was prepared by JSSC [53].These days, however, some practical examples can be seen in some connections of steel bridge members. Above all, the long connection type of joints is often used.However, the short connection type is only used for connections of such temporarystructures as erection facilities and so on. The reason why is that enough information

about the short connection type to apply them to primary members of actual steelbridges is lacking even in the recommendation.Some examples of long connection type joints in Japan are shown in Fig. 48. Fig.

48(a) shows a connection of a tower of three suspension bridges, Kurushima Bridges[54]. These are continuous suspension bridges constructed in 1999. The height of one tower of the longest bridge is about 180 m. This is the rst example of the longconnection type joint in Japan. In this case, predominant cross sectional force at the joint of the tower is axially compression, and tensile force induced in some boltsdue to bending moment is not predominant. By using this type of joints, no erectionfacilities outside the towers are needed. As a result, easiness and safety of the erection

can be substantially improved. Fig. 48(b) shows bridge restrainers using wire cablesfor preventing elevated girders from falling against strong earthquake. Long connec-tion type joints are also adopted for anchoring the cables. Fig. 48(c) shows the anchor


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F i g

. 4 7

.

T w o t y p e s o f t e n s i l e j o i n t s

. ( a ) S h o r t c o n n e c t i o n t y p e ; ( b ) L o n g c o n n e c t i o n t y p e .


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Table 6Features of tensile joints

Short connection type Long connection type

Construction Very easy Fairly easyConnecting tools High strength bolts for friction PC tendons

type jointsPrying force Not negligible NegligibleDuctility Smaller Larger

bolts supporting a reaction force wall of a big oating bridge in Osaka. Theseexamples have been designed according to the JSSC recommendation.

A typical example of the short connection type joint is shown in Fig. 49. Theshort connection type joint connecting a main girder and a crossbeam using T shapegussets can be seen in this photo. These joints are sometimes used in recent bridgesin Japan, which have two or three main plate girders and thick PC slabs of highdurability. The thickness of the ange plates and web plates of the main girders is34 mm and about 14 19mm, respectively. The design of the joints is also based onthe JSSC recommendation.

In addition, other potential examples in which tensile joints can be properly appli-cable are also illustrated in Fig. 50. Ease and safety of construction can be expectedin these examples. These examples will be realized in the future by overcoming

technical issues as mentioned later.5.2.3. Concept of design procedure

The JSSC recommendation de nes that the ultimate limit state of the tensile jointis decided by fracture of high strength bolts. The reason why the ultimate limit stateof the joint is de ned by the fracture of the bolt in the JSSC recommendation is touse the high strength of bolts effectively, and to simplify its design method. Basicdesign concepts of the short/long connection types are summarized in Table 7,respectively.

Connecting tools for the short connection type joint are same high strength bolts

as for the friction type joint, and for the long connection type used are PC tendons.In particular, by taking constructability into consideration, the same high strengthbolts as for the friction type joints are also used for the short connection type joint,and the axial force introduced into the bolts is almost the same as that of the frictiontype joint.

The evaluation of the prying force is the most important factor in the design of the short connection type joints. A lot of evaluation formulas have been proposeduntil now; some of them are used in design speci cations for tensile joints in theworld. Evaluation formulas used in them are summarized in the JSSC recommen-dation. The JSSC recommendation uses the Kato Tanaka formula [55]. On the other

hand, the recommendation for building structures in Japan [56] adopts the Tanaka Tanaka formula [57]. These formulas are classi ed into two categories according tothe applied position of the prying force; one assumes that the prying force is applied


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Fig. 48. Example of long connection type joints. (a) Horizontal joint of a suspension bridge tower,Kurushima Bridges; (b) restrainers of connecting girders against earthquake; (c) anchor bolts for a reactionwall of a oating bridge under erection in Osaka.


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Fig. 49. Example of short connection type joints.

Fig. 50. Expected examples of tensile joints in box girder bridges. (a) Joints of box cross sectionalmembers; (b) Examples of connected section.

to an edge of the ange plate, and the other one assumes that the applied position

is actually decided by the ultimate deformation mode of the ange plate. Theseformulas are derived through using cantilever beam models as shown in Fig. 51.Each formula is mainly based on the experimental results on beam to column connec-


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Table 7Design concepts of tensile joints

(a) Short connection typeAssumed ultimate limit state Breaking of a boltConnecting tools High strength bolts for friction type (F10T and

S10T are standard)Introduced axial force Same as that of friction typeBolt arrangements Same as those of friction type (only 1 row)Minimum thickness of ange plate More than nominal diameter of high strength bolt

usedLimitation of connected members Suf cient stiffening is required, and the minimum

thickness is more than 1/2Calculation of prying force Kato Tanaka formula(b) Long connection type

Assumed ultimate limit state Breaking of a boltConnecting tools PC tendons (material is equivalent to F10T)Introduced axial force Same as that of friction typeMinimum thickness of anchor plate 22 mm (bearing plate type); 30 mm (except

bearing plate type)Width of rib plates used in stiffened sections About 5 times of bolt diameter used

Fig. 51. Prying force evaluation model in JSSC recommendation.


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tions in the past. In the JSSC recommendation, therefore, applicability of the Kato Tanaka formula has been investigated considering the dimensions of bridge structuresthrough a nite element 2D-analysis. The design ow diagram for the short connec-tion type joints according to the JSSC recommendation are drawn in Fig. 52. Thedifference between these ow diagrams is whether the additional bolt axial forcedue to the prying action is considered.

5.2.4. Issues to be improved in Japanese recommendations for tensile jointsIn Japan recently, some steel bridges are made of comparatively thick steel plates

with less numbers of stiffeners in order to reduce fabrication cost. Since it is speci edin JSHB [36] that the allowable strength of the connections should be higher thanboth maximum stress due to applied loads and 75% of the full allowable strengthof the stronger one of the connected sections, some connections designed by theJSHB tends to be conservative. This requires a lot of bolts consequently. In the JSSCrecommendation, however, there is no information on multiple arrangements of bolts

Fig. 52. Design ow diagram of tensile joints, short connection type (JSSC recommendation).


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like the friction type joint as illustrated in Fig. 53. In addition, dimensions of jointdetails, such as bolt arrangement are still based on that of the friction type jointbecause of lack of enough information on the mechanical behavior and avoidanceof confusion between the tensile type joint and the friction type one. The rationalbolt arrangement of tensile joints for the use of much higher performance is, there-fore, going to be speci ed considering the difference of load transferring mechanismbetween the tensile and friction type joints.

The material of the connecting tools, such as high strength bolts and PC tendons,must be also improved in order to realize high performance tensile joints, namely,high strength and high ductility ones. And the design speci cations for them arealso needed.

Moreover, in the case where these tensile joints are applied to primary membersof bridges, the rigidity of the connections is very important. However, there is notenough information on them in the JSSC recommendation. Accordingly, rst of all,the de nition of the rigidity of the tensile joints should be made clear and then theeffect of the rigidity on the mechanical behavior of the stiffness and strength of theconnected members should be investigated in the future.

5.3. Recent research activities on tensile joints in Japan

Various studies on high strength bolted tensile joints are carried out in Japan, inorder to provide and accumulate basic information needed for applying this type of

joints to connections of primary members in bridges and then for revising the JSSCrecommendations of tensile joints. Studies on tensile joints by authors and the latesttopics in this eld in Japan are brie y introduced as follows:

5.3.1. High ductility boltsThe effect of the arrangement of the bolts in multiple bolt lines for increasing the

tensile strength of the joints is not so high compared with the multiple bolt arrange-ments of friction type joints. This may be caused by less ductility of ordinary high

Fig. 53. Schematic view of multiple bolt arrangements of tensile joints. (a) One row bolt arrangement;(b) 2 rows bolt arrangement.


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Fig. 54. High ductility bolt.

strength bolts. It is, therefore, considered that the ultimate strength of the tensile joint with multiple bolt lines can be improved by using a waisted shank bolt asshown in Fig. 54. Similar bolts to this type of bolts are speci ed in BS [58]. Theenough diameter of the waisted shank to increase the strength of the joints is almosta little smaller one than the effective diameter of ordinary bolts [59].

Fig. 55 shows the effectiveness of the waisted shank bolt for tensile joints. It canbe seen that the yield strength of the joint substantially increases in particular, whilethe ultimate strength increases a little. It can be recognized that the outer bolts areable to transfer about 30% of the applied load by using waisted shank bolts at theinner side.

5.3.2. High strength bolted tensile joints for box cross sectionsExperimentally investigated in Ref. [60] is the applicability of tensile joints to

connections of main bridge members with box cross-section. The tested specimensas shown in Fig. 56 are to investigate the behavior of box members connected bythe tensile joints. The nominal diameter of the bolts (M12) is 12 mm. In the experi-ments, bending moment and tensile force are applied simultaneously to the speci-mens. Attention is paid to the effect of the bolt arrangements, rigidity of the joints,

Fig. 55. Effectiveness of high ductility bolts.


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Fig. 56. Box cross sections of seven specimens with variation of bolt arrangements.

and increase of the bolt axial force. Bending moment vs. curvature curves aredepicted in Fig. 57. The horizontal axis of the gure denotes the average curvatureof the jointed section, and the vertical axis is the applied bending moment. The yield

point is de ned as yielding of the severest one of the bolts. And the yield points of the box cross-sections without any joints are also drawn in this gure. It can beunderstood that the thicker the end plate is, the higher the ultimate strength becomes.However, the ultimate strength of the specimen with the waisted shank bolt doesnot increase, but the ductility of this tensile joint is improved. The failure modesdeciding the ultimate limit state of the tensile joints can change from the brittle oneto the ductile one by using waisted shank bolts. Comparing A34 with D34, it canbe also recognized that the arrangement of the outer row bolts is very effective inthe aspects of high strength, high rigidity and high ductility. It can be seen from therelationships between the yield strength and the ultimate strength points of all the

specimens that the ultimate strength are almost at least 1.5 times the yield strength.The tensile joint becomes, therefore, very ductile up to the ultimate limit state, if the joint is designed based on the yield strength. According to these results mentionedabove, it can be recognized that the high strength bolted tensile joint for box-crosssections are applicable to real bridge structures.

5.3.3. Super high strength boltsA high strength bolt, of which tensile strength is 1400 MPa, has been newly

developed [61, 62]. In Japan, high strength bolts with tensile strength higher than1000 MPa have not been made and used because of delayed failure since about 30

years ago. Usage of bolts with much higher strength is most effective for strengthen-ing the joints, especially tensile joints, although further investigation is necessary asfar as fatigue strength and delayed failure is concerned. The shape of the bolt should


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Fig. 57. Bending moment vs. Curvature curves.

be, therefore, determined so as to be able to avoid high stress concentration near thethreaded portion. An example of the bolt is shown in Fig. 58. However, there is nopractical joint using those bolts for bridge structures for lack of experimental resultson the fatigue strength and delayed failure.

In Japan, usage of high strength bolted joints for bridge structures has been limited.

Fig. 58. Example of super high strength bolt (unit : mm).


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Much more usage of such joints can be expected by overcoming issues, such asproper bolt arrangements for the joints, adoption of new material, rational evaluationof ultimate strength, namely slip strength for friction type joints and ultimate strengthfor tensile type joints considering prying action force and joint rigidity in the future.And, in order to make bridge structures more rational both in economical aspect andstructural performance, the high strength bolted joints, especially tensile joints, willbe one of promising methods for led connections.

6. Conclusions

This paper has introduced computer programs developed mainly by authors for

advanced static/dynamic elasto plastic nite displacement analyses of steel bridgestructures, the ultimate strength and design methods of steel plates and columnsmade of high strength steel subjected to compression, seismic design and retro ttingmethods of new and existing steel bridge piers, and friction type joints and tensiontype joints with high strength bolts and high performance high strength bolts inJapan, among new technologies of steel bridges under development in Japan.

In the paper, the part on the computer programs for advanced static/dynamic ela-stoplastic nite displacement analyses were described by Kitada and Okada. Theultimate strength and design methods of steel plates and columns made of highstrength steel were mainly written by Ochi. The part of the seismic design and retro-

tting methods of steel bridge piers were by Ono and Matsumura, respectively. Jointswith high strength bolts were described by Yamaguchi.

Acknowledgements

The authors wish to thank Vice-Chancellor P.J. Dowling of Surry University andProfessor S. Nakamura of Tokai University for giving us a good chance to introduceour researches in Japan to the world through this journal.

References

[1] Subcommittee on Common Design Code on Steel and Concrete Structures, JSCE (Japan Society of Civil Engineers). General Principles for the Limit State Design of Steel and Concrete Structures.In: Proceedings of JSCE, No.450/I-20. July 1992:13 20 [in Japanese].

[2] Subcommittee on Design Code for Steel Structures, Committee on Steel Structures, JSCE. DesignCode for Steel Structures, Part A. In: Nishino F, editor. Structures in general, April 1997 [in Japane-se].

[3] Subcommittee on Ultimate Strength of Steel Structures and their Task Group of Ultimate Strengthof Composite Structures, Committee on Steel Structures, JSCE. Design Code for Steel Structures,Part B. In: Composite structures, May 1997 [in Japanese].

[4] Kitada T, Ohminami R, Nibu M, Tanaka K. Exploitation of computer program to analyze ultimatestrength of steel bridges using cable members. In: Proceedings of Symposium on Computational


48/50


Methods in Structural Engineering and Related Fields, Japanese Society of Steel Construction(JSSC), vol. 13, July 1989:89 94 [in Japanese].

[5] Kitada T, Ohminami R, Nibu M, Kano M, Tanaka K. Effective approach for elasto plastic and nite

displacement analysis of steel bridges with cables. In: Proceedings of Symposium on ComputationalMethods in Structural Engineering and Related Fields, JSSC, vol. 14, July 1990:299 302 [in Japane-se].

[6] Kitada T, Ohminami R, Nibu M, Tanaka K, Kano M. On various strategies for elasto plastic andnite displacement analysis of three dimensional framed structures. In: Proceedings of Symposiumon Computational Methods in Structural Engineering and Related Fields, JSSC, vol. 14, July1991:237 42 [in Japanese].

[7] Kano M, Kitada T, Nibu M, Tanaka K. Effective strategies for elasto plastic and nite displacementanalysis of spatial steel bridges. In: Proceedings of (SSRC) Structural Stability Research Council,Milwaukee (USA), April 1993:333 43.

[8] Tanaka K, Kitada T, Nibu M, Kano M. On a computer program, EPASS, to analyze ultimate loadcarrying capacity of spatial steel bridges structures. In: Proceedings of SSRC, Milwaukee (USA),

April 1993:357 67.[9] Kitada T, Nakai H, Tanaka K. Fundamental theory of computer program, EPASS, for applyingelasto plastic and nite displacement analysis to spatial framed steel bridges. In: Memoirs of theFaculty of Engineering, Osaka City University, vol. 39, December 1998:53 69.

[10] Tanaka K, Kitada T, Mukaiyama T. Elasto plastic nite displacement analysis of box membersmode of steel plates with no yield plateau and tapered box members. Journal of Structural Engineer-ing, JSCE 2000;46A:271 9.

[11] Kitada T, Nakai H, Yoshikawa O, Sakano M. On a design method against buckling stability of archribs of Nielsen Lohse bridges. Journal of Structural Engineering, JSCE 1988;34A:315 26.

[12] Kitada T, Nakai H, Ashihara E, Tanaka K. Elasto plastic nite displacement behavior and ultimatestrength of large or long-span steel bridges. In: Proceedings of 5th East Asia-Paci c Conference onStructural Engineering and Construction (EACEC5), Queensland (Australia), July 1995:279 84.

[13] Investigation & Research Committee on Road Bridges, Kansai Road Study Association. Ultimatestrength analysis and buckling design of steel bridges by computer. Kyoritsu Shuppan Book Com-pany, February 1998 (in Japanese).

[14] Tanaka K, Kitada T. On elasto plastic and nite displacement analysis for calculating load carryingcapacity of Steel bridge structures. Journal of Structural Engineering, JSCE 2001;47A:125 36.

[15] Noguchi J, Kitada T, Hikiguchi M. Relationships between safety factors of structural members andoverall load carrying capacity of steel cable-stayed bridges. JSCE 2001;47A:137 46.

[16] Sakoda H, Zui H, Kagayama T, Uehira S, Kitada T, Nakai H. On causes of loosening and slacknessof hanger cables in a Nielsen Lohse bridge due to the Hyogo-ken Nambu Earthquake. Steel Con-struction Engineering, JSSC 2000;7(25):31 42.

[17] Kitada T, Sakoda H, Tanaka K, Kagayama T. Causes of buckling damage to upper lateral bracingmembers in a large Lohse arch bridge due to the Hyogo-ken Nambu Earthquake. Journal of StructuralEngineering, JSCE 2000;46A:841 50.

[18] Kitada T. On seismic design method for steel bridges under large earthquake based on elasto plastic nite displacement analysis. In: Proceedings of 5th International Colloquium on Stability andDuctility of Steel Structures, vol. 2, Nagoya, Japan, July 1997:763 70.

[19] Komatsu S, Kitada T, Miyazaki S. Elasto plastic analysis of compressed plates with residual stressand initial de ection. In: Proceedings of JSCE, No. 244, December 1975:1 14 [in Japanese].

[20] Komatsu S, Kitada T. Elasto plastic nite displacement analysis of longitudinally stiffened platesin compression. In: Proceedings of JSCE, No. 296, April 1980:1 12 [in Japanese].

[21] Kano M, Yamano T, Nibu M, Kitada T. A computer program, USSP, for analyzing ultimate strengthof steel plated structures. In: Proceedings of 5th International Colloquium on Stability and Ductilityof Steel Structures, vol. 2, Nagoya, Japan, July 1997.

[22] Kitada T. Section 5.3.2 stiffened plates under biaxial compression. In: Fukumoto Y, editor. Structuralstability design. Pergamon, UK, 1997:162 67.

[23] Nakai H, Kitada T, Iwai Y. Parametric analysis on ultimate strength of compressive stiffened plateswith opening. Journal of Constructional Steel, JSSC 1999;5:231 8.


49/50


[24] Nakai H, Kitada T, Murayama Y, Murozuka N. An analytical study on ultimate strength of boxgirders subjected to bending and torsion. Journal of Structural Engineering, JSCE 1996;42A:71 82.

[25] Nakai H, Kitada T, Kunihiro M, Harada N, Oryu T. Analytical study on interaction curve for ultimate

strength of thin-walled box stub columns. Journal of Structural Mechanics and Earthquake Engineer-ing, JSCE 1998;605/I-45:265 76.[26] Jun TJ, Nakai H, Kitada T, Kano M, Tanaka T. A study on lateral torsional buckling and ultimate

strength of horizontally curved I-girders by elasto plastic and nite displacement analysis. In: Pro-ceedings of 5th Korea Japan Joint Seminar on Steel Bridges, Pusan, Korea, October 1999:479 90.

[27] Kitada T, Nakai H, Ochi N. Study on ultimate strength of plates and stiffened plates made of highstrength steel under compression. Journal of Structural Engineering, JSCE 2000;46A:179 90.

[28] Kitada T, Nibu M, Kano M, Yamano T. Elasto plastic and nite displacement analysis consideringinteraction between local and overall bucklings of thin-walled framed structures by nite elementmethod. In: Proceedings of the Conference on Computational Engineering and Science, vol. 2, No.3,the Japan Society for Computational Engineering and Science (JSCES), March 1997:925 28 (inJapanese).

[29] Kano M, Nibu M, Kitada T. Elasto plastic and nite displacement analysis of thin-walled framedstructures by nite element method. In: Proceedings of 7th KAIST-NTU-KU TrilateralSeminar/Workshop on Civil Engineering, Kyoto, Japan, December 1997:75 80.

[30] Kitada T, Nibu M, Kano M, Yamano T. Interactive analysis between local and overall buckling of thin-walled framed structures. Theoretical and Applied Mechanics 1998;47:47 55.

[31] Mizukoshi M. Elasto plastic analysis of thin-walled framed structures considering interaction of local and overall bucklings. Master thesis, Osaka City University, March 1999 (in Japanese).

[32] Kitada T, Nakai H, Kano M, Okada J. Establishment of a method for analyzing elasto plastic anddynamic response of steel bridge piers with single columns by considering local buckling. In: Pro-ceedings of the Second Symposium on Nonlinear Numerical Analysis and its Application to SeismicDesign of Steel Structures, November 1998:255 62 (in Japanese).

[33] Committee for Developing USSP. USSP (Ultimate Strength of Steel Plated Structures), User s Man-ual (Basic Theory), Ver. 3.0, Bridge and Computer Engineering Co., Ltd. October 1996 [in Japanese].

[34] Nakai H, Kitada T, Nakanishi K. Hybrid test for simulating seismic behavior of steel and compositebridge piers. Memoirs of the Faculty of Engineering, Osaka City University, vol. 37, December1996:49 60.

[35] Ministry of Construction, PWRI and 5 other Organizations. Ultimate limit state design method of highway bridge piers under seismic loading. Joint Research Program Report, vol. 7, April 1997.

[36] Japan Road Association. Speci cations for highway bridges, part II. Steel bridges, Maruzen,December 1996 [in Japanese].

[37] Komatsu S, Ushio M, Kitada T. An experimental study on residual stress and initial deformationsof stiffened plates. In: Proceedings of JSCE, No. 265, September 1977:25 35 [in Japanese].

[38] Komatsu S, Kitada T. Ultimate strength characteristics of outstanding steel plates with initial imper-

fections under compression. In: Proceedings of JSCE, No. 314, October 1981:15-27 [in Japanese].[39] Honshu-Shikoku Bridge Authority. The design speci cations for superstructures, April 1989 [in

Japanese].[40] Seismic Design WG, Committee on New Technology for Steel Structures. A proposal for seismic

design of steel bridges, Japan Society of Civil Engineers (JSCE), 1996 [in Japanese].[41] Japan Road Association. Japanese speci cations for highway bridges, part V: seismic Design,

December 1996 [in Japanese].[42] Nishikawa K, Murakoshi J. Seismic design for steel bridge piers. The Foundation Engineering and

Equipment 1997;25(3).[43] Ministry of Construction. Report on the damage to highway bridges by the Hyogo-ken Nambu

Earthquake, Committee for Investigation on the Damage to Highway Bridges Caused by the Hyogo-ken Nambu Earthquake, 1995 [in Japanese].

[44] Public Works Research Institute of the Ministry of Construction, Metropolitan Expressway PublicCorporation, Hanshin Expressway Public Corporation, Nagoya Expressway Pubic Corporation, JapanAssociation of Steel Bridge Construction and Kozai Club. Ultimate limit state design method for


50/50


highway bridge piers under seismic loading, cooperative research report, No.178 184, 212, 219[in Japanese].

[45] Hanshin Expressway Public Corporation. Guidelines for seismic retro tting methods for existing

steel bridge piers (Draft), March 1998 [in Japanese][46] Kitada T, Nakai H, Kagayama T, Matsumura M. Study on seismic retro tting method of stiffenedplates in existing bridge piers. Journal of Structural Engineering, JSCE 1999;45:215 26.

[47] Kitada T, Nakai H, Matsumura M, Kagayama T. Experimental study on seismic retro tting methodof stiffened plates in existing steel bridge piers under cyclic loading. Journal of Structural Engineer-ing, JSCE 2000;46:127 34.

[48] Akiyama H, Nishiura N, Kamei Y, Ikehata B. Classi cation of limit states on HSFG bolted jointin tension. Journal of Constructional Steel, JSSC 1995;3:289 96.

[49] Akiyama H, Nishimura N. Slip mechanism and evaluation on limit strength of HSFG bolted jointin plate-girder under bending. Journal of Constructional Steel, JSSC 1996;4:287 94.

[50] Japanese Society of Steel Construction (JSSC). Design manual for rational steel bridges in economi-cal and mechanical aspects, JSSC, 2000 [in Japanese] pp. 84 113.

[51] Kamei K, Nakamura T, Akiyama H, Takagi M, Nishimura N. An experimental study on slip

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