FOUNDATION DESIGN

University of Nairobi –FCE 511 Geotechnical Engineering IV

University of Nairobi

Department of Civil and Construction Engineering

Geotechnical Engineering (FCE 511)

Teaching notes

By Sixtus Kinyua Mwea

2015

University of Nairobi –FCE 511 Geotechnical Engineering IV i

Syllabus

FCE 511 - Geotechnical Engineering III

Foundations:

Shallow Foundations

Introduction. Foundation loading intensities. Bearing capacity, (ultimate, safe, gross and

allowable). Influence of ground water table, sloping ground, inclined and eccentric loads on

allowable bearing capacity. Design of shallow foundations for shear strength and settlements.

Examples of foundation design (e.g. strips, pad), combined footings, raft footings.

Piled Foundation

Types of piles driven and bored pile, friction and end bearing pile. Design of piles by soil

mechanics methods, end bearing, skin friction and ultimate bearing resistance. Piles in sands.

Piles in cohesive soils - total and effective stress analysis. Design from pile tests data.

End bearing piles on rock. Settlement of piles. Dynamic formula. Negative skin friction. Pile

groups - bearing capacity in cohesive and cohesionless soils.

Introduction to Earth Dams

Design of earth embankment - homogenous and zoned dams. Definitions e.g. fetch, water

spread, shell free board etc. Factors influencing site selection. Spillways. Settlements of

embankments. Protection of upstream and downstream slopes.

Site Investigations

Introduction, purpose of Site Investigation, organization of Site investigation for different types

of structures e.g. buildings, irrigation or water supply projects, highways and airport pavements,

etc. Methods of Investigation. Sampling. Borehole logs. Geophysical methods. Geotechnical

reports.

University of Nairobi –FCE 511 Geotechnical Engineering IV ii

Table of contents

Syllabus ................................................................................................................................... i

Chapter one ................................................................................................................................ 1

Shallow foundations ................................................................................................................... 1

1.1 Types of foundations .................................................................................................. 1

1.2 Introduction to shallow foundations ........................................................................... 1

1.2 Bearing capacity of soils ............................................................................................ 2

1.2.1 Bearing capacity terms ......................................................................................... 2

1.2.2 Ultimate bearing capacity ..................................................................................... 3

1.2.3 The net foundation pressure ............................................................................... 12

1.2.4 Allowable bearing pressure ................................................................................ 13

1.2.5 Field methods for the determination of bearing capacity of soils ...................... 14

1.2.6 Presumed bearing capacity of soils and rocks .................................................... 23

1.3 Proportioning of shallow foundations ...................................................................... 24

1.3.1 Contact pressure distribution .............................................................................. 24

1.3.1 Proportioning the foundations ............................................................................ 25

1.3.2 General consideration in the selection of the foundation depth ......................... 34

1.3.3 Foundations for common buildings .................................................................... 35

1.4 Foundations for difficult soils .................................................................................. 36

1.4.1 Foundations on expansive clays ......................................................................... 36

1.4.2 Foundations on loose sands ................................................................................ 41

1.5 Tutorial examples on chapter one ............................................................................ 43

Chapter two .............................................................................................................................. 45

Deep Foundations ..................................................................................................................... 45

2.1 Pile foundations ........................................................................................................ 45

2.1.1 Introduction ........................................................................................................ 45

2.1.2 Classification of Piles by materials and construction ......................................... 46

2.1.3 Driven piles ........................................................................................................ 48

2.1.4 Bored piles .......................................................................................................... 51

2.1.5 Determination of pile load carrying capacity ..................................................... 53

2.1.6 Determination of load carrying capacity dynamic methods ............................... 59

2.1.6 Determination of load carrying capacity pile testing ......................................... 61

2.1.7 Negative skin friction ......................................................................................... 62

University of Nairobi –FCE 511 Geotechnical Engineering IV iii

2.1.8 Pile groups .......................................................................................................... 64

2.2 Drilled piers and Caisson Foundations..................................................................... 66

2.2.1 Drilled piers ........................................................................................................ 66

2.2.2 Caisson Foundations .......................................................................................... 66

2.4 Examples of Piling Schemes .................................................................................... 71

2.5 Tutorial examples on chapter two ............................................................................ 71

Chapter Three ........................................................................................................................... 73

Introduction to Earth Dams ...................................................................................................... 73

3.1 Introduction .............................................................................................................. 73

3.2 Selection of type of earth dam .................................................................................. 74

3.2.1 Diaphragm types ................................................................................................ 74

3.2.2 Homogenous types ............................................................................................. 75

Zoned types ..................................................................................................................... 75

3.2 Design Principles ..................................................................................................... 76

3.3.1 Foundation design .............................................................................................. 76

3.3.2 Embankment Design .......................................................................................... 79

3.3 Inspection of existing dams ...................................................................................... 81

3.4 Examples of earth dams in Kenya ............................................................................ 82

Chapter Four ............................................................................................................................. 88

Site Investigation ...................................................................................................................... 88

4.1 Introduction .............................................................................................................. 88

4.1.2 Planning a site investigation ............................................................................... 89

4.2 Preliminary and detailed stage site investigations .................................................... 91

4.2.1 Preliminary stage site investigations .................................................................. 91

4.2.2 Detailed stage site investigations ....................................................................... 92

4.2.3 Sampling ............................................................................................................. 97

4.2.4 Scope of Site Investigation ............................................................................... 101

4.2.5 Site Investigation Reports ................................................................................ 102

References: ............................................................................................................................. 103

University of Nairobi –FCE 511 Geotechnical Engineering IV iv

Geotechnical Engineering IV

Week 1 2 3 4 5 6 7 8 9 10 11 12 13 14 15

Introduction

Shallow foundations

Foundation intensities

Bearing Capacity

Factors that influece bearing capacity

Design of shallow foundations

Piled foundations

Types of piles

Types of piles

Driven piles

Bored piles

Pile load capacity

Settlement of piles

Negative skin friction

Pile groups

Introduction to earth dams

Definitions (fetch water spread,

freeboard)

Design of earth embankment

Site selection

Spillways

Settlement of embankments

Protection of slopes

Continuous Assessment Test

Site investigation

Introduction

Purpose of site investigation

Organization of site investigation

SI for different schemes

Methods for site investigation

Geotechnical reports

Revision and tutorials

Main examinations

Target dates

University of Nairobi –FCE 511 Geotechnical Engineering IV -1-

Chapter one: - Shallow foundations

1.1 Types of foundations

Foundations that are encountered in practice may be classified into two broad categories

namely shallow and deep foundations. Under shallow foundations the following categories

are usually encountered:-

a) Strip foundations for wall and closely spaced columns

b) Spread or isolated footings for individual columns. In this category it is usual

to consider combined foundations for two or three closely spaced columns as

spread or isolated footings

c) Raft foundations covering large sections of the foundation area

The design and construction of shallow foundations is dealt with in this chapter.

Under deep foundations the following two types of foundations are encountered :-

a) Piles

b) Caissons

The design and construction of deep foundations is dealt with in the next chapter.

In the selection of the foundations to adopt for a structure it is usually necessary to

consider the function of the structure, its loads, the subsurface conditions and the cost of the

foundation being adopted in comparison to other possible types of foundations.

1.2 Introduction to shallow foundations

The foundation is the part of the structure that transmits the loads directly to the underlying

soil. If the soil is sufficiently strong it is possible to use shallow foundation. On the other

hand if the soil is not strong enough the foundation is taken deeper into the ground and is

referred to as a deep foundation. A definition which sometimes conflicts with the definition

of the shallow foundation defines a shallow foundation as one whose depth is less or equal to

its least width. The foundation must satisfy two fundamental requirements:-

- 2- Shallow foundations


1. The factor of safety against shear failure must be adequate. A value of 3 to 5 is

usually specified.

2. The settlement of the foundation should be tolerable and in particular differential

settlement should not cause any unacceptable damage o interfere with the function of

the structure.

3. The allowable bearing capacity is defined as the pressure which may be applied to the

soil to enable the two fundamental conditions to be satisfied

The damage being mitigated in the design of the structures can be classified as architectural,

functional or structural. In the case of framed structures settlement damage is usually

confined to the cladding and finishes (architectural damage). It is usual to expect a certain

amount of damage. What is critical is to ensure that the damage to the services is limited.

Angular distortion limits were proposed by Craig (1987) and are shown on Table 1.1. In

general the limiting angular distortion to prevent damage is 1/300. For individual footings

this translates to a maximum settlement of about 50mm in sand and 75mm in clay. An

accurate damage criterion is to limit the tensile strain at which the cracking occurs. The

concept of tensile strain should be used in analysis using an idealization of the structure and

the foundation in elastic strain analysis when the fundamental properties of the foundations

are known.

Table 1.1 Angular distortion limits

1/150 Structural damage of general buildings may be expected

1/250 Tilting of high rigid buildings may be visible

1/300 Cracks in panel walls expected

Difficulties with overhead cranes

1/500 Limit for buildings in which cracking is not permissible

1/600 Overstressing of structural frames without diagonals

1/750 Difficulties with machinery sensitive to settlement

The design of the foundations is usually a two process exercise. The first is to determine the

allowable bearing of the soil while the second is to size the foundation on the design strata

based on the allowable bearing capacity. These two parts are now discussed.

1.2 Bearing capacity of soils

1.2.1 Bearing capacity terms

The following terms are used in bearing capacity problems

Ultimate bearing capacity is the value of the average contact pressure between the foundation

and the soil which will produce shear failure in the soil.



The net foundation pressure is the increase in the pressure at the foundation level due to the

structure loads

The safe net foundation pressure is the net foundation pressure divided by a suitable factor of

safety

Allowable bearing pressure is the maximum allowable net loading intensity on the soil

allowing for both shear and settlement effects.

1.2.2 Ultimate bearing capacity

If a load is increased at the foundation level, shear failure would take place in the foundation

at a load which can be referred to as failure load. The resulting pressure at the base of the

foundation is known as the ultimate bearing capacity of the soil

Three distinct modes of failure have been identified and these are illustrated in

Figure 1.1 in the case of strip footing. As the pressure increases on the foundation layer the

state of plastic equilibrium is reached initially in the soil around the edges of the footing and

then spreads downwards and outwards. Ultimately the state of plastic equilibrium is reached

throughout above the failure surfaces. The soil around the footing heaves on both sides. At

the moment of failure one side continues to settle at a higher rate and the strip footing tilts.

This behavior is exhibited by soils of low compressibility (Figure 1.1a). . Local shear failure

is characterized by local development of plastic conditions usually below the foundation. The

plastic conditions do not reach the surface and only slight heaving is expected. This kind of

failure is expected with soils of high compressibility and is associated with large settlements

(Figure 1.1b). These soils include dense and stiff soils. Punching shear occurs when shearing

takes place directly below the footing under compression from load. No heaving is of the

ground is expected by the side of the footing. Large settlements are characteristics of this

mode of failure and are typical of soils of high compressibility and foundations at

considerable depth (Figure 1.1c). In general the mode of failure will depend of the

compressibility of the soil and the depth of the foundation.



a) General shear failure

b) local shear failure

c) Punching shear failure

Figure 1.1 Modes of failure of foundations

Bearing capacity by use of earth pressure analogy

The earth pressure analogy can be explained by consideration of a strip footing on a

cohesionless soil as shown on Figure 1.3

Figure 1. 2 Pressure below a strip footing

The vertical pressure is q which is a result of the structure loads. By use of Rankine active

pressure theory, a lateral pressure p holds the soil in equilibrium below the foundation. For

particles just beyond the edge of the foundation the lateral pressure is more than the vertical

pressure γD resulting from the overburden. The vertical pressure γD is the minor principle

stress and p is the principal stress. By use of the Rankine earth pressure theory Equations 1.1

through 1.3 can be deduced.

)sin1/()sin1( qp (inside the foundation) 1. 1

)sin1/()sin1( Dp (outside the foundation) 1. 2

2))sin1/()sin1(( Dq (ultimate bearing capacity) 1. 3

q

γD p

c b

a

Set

tlem

ent

Pressure



For a cohesionless soil the bearing capacity is dependent on the overburden and equals to

zero for a foundation on the ground surface. Bells development for a c-υ is given in Equation

1.4

)sin1/()sin1(2)sin1/()sin1((2))sin1/()1(( 32 ccsiDq .1. 4

For a purely υ =0 soil the ultimate bearing capacity is given by Equation 1.5

cDq 4 1. 5

Bearing capacity by use slip circle analogy

The slip circle analogy can be explained by consideration of a strip footing on a cohesive soil

as shown on Figure 1.3

Figure 1. 3 A slip circle analogy of a strip footing

The foundation is assumed to fail by rotation about a slip surface of radius equal to the width

of the base B and at the edge of the foundation O. At ultimate conditions the disturbing

moment (Md) is given by Equation 1.6

2***

BBLqMd 1. 6

The resisting moment (Mr) about O is a summation of the resistance due to the cohesion on

the cylindrical surface, on the vertical surface and the weight of the overburden as given in

Equation 1.7

2

22 DLB

CDLBcLBMr 1. 7

At ultimate conditions the disturbing moment is equal to the resisting moment and the

ultimate bearing Equation for a υ = 0 soil is given by Equation 1.8

B

O

πB

q

B

D



)16.032.0

1(28.6cD

BD

cq

1. 8

Plastic theory failure

A suitable failure under a strip footing is shown on Figure 1.2. The footing of width b and

infinite length carries a uniform pressure of magnitude qf. The shear strength parameters for

the soil are c and υ. The unit weight of the soil is assumed to be zero. At ultimate bearing

capacity the soil is pushed downwards into the soil mass producing a state of plastic

equilibrium in the form of an active Rankine zone below the footing where the angles ABC

and BAC are each 45+υ/2. The zone ABC resists movement and is intact with the base. It

suffers no much deformation. The downward movement of the wedge ABC forces the

adjoining soil to move sideways. Passive Rankine zones ADE and GBF are developed and

angles AEF and BFG are 45-υ/2. these zones confine the movement of the wedge EDA and

BGF. The transition between the downward movement of the wedge ABC and the lateral

movement of the wedge EDA and BGF takes place through zones of radial shear ACD and

BCG. The surfaces DC and CG are logarithmic spirals. The soil above EDCGF is in a state

of plastic equilibrium while the rest of the soil is in state of elastic equilibrium.

Figure 1. 4 Failure under a strip footing

Using plastic theory the ultimate bearing capacity below a strip footing on a surface of a

weightless soil is given by Equation 1.9. This is for undrained condition where υu = 0

uuf ccq 14.5)2( 1.9

In general the foundation is located at a depth and imposes a surcharge qo = γD. The weight

of the surcharge and the pressure of the foundation produce stresses on the moving masses of

soil at plastic conditions.

The ultimate bearing capacity of the soil under shallow strip footing can be

expressed by the following general equation suggested by Terzaghi.

qcf DNCNBNq 5.0 1. 10

qo

qf

45-φ/2

45+φ/2 A B

D C

E

G

F



Nγ, Nc and Nq are bearing capacity factors which depend on the values of υ. Nγ

represents the contribution to the bearing capacity resulting from the self weight of the soil.

Nc is the contribution due to the constant component of the shear strength and Nq is the

contribution of the surcharge pressure. Values of Nγ, Nc and Nq can be obtained from

Equations 1.11 through 1.13 the values for Nc and Nq were suggested by Meyerhof (1955)

while the values of Nγ, were suggested by Hansen (1970) These values are plotted in terms of

υ in Figure 1. 5.

cot)1( qc NN 1. 11

tan2 )2/45(tan eNq 1. 12

tan)1(5.1 qNN 1. 13

Figure 1. 5 Bearing capacity factors for shallow foundations

Bearing capacity for square, round and rectangular foundations

The problem involves extending what is basically a two dimension problem in a strip footing

to a three dimension problem in other foundation shapes. The bearing capacity factors for

square and round foundations are shown on Equations 1.14 and 1.15 respectively.

qc DNcNBNq 3.14.0 1. 14

qc DNcNBNq 3.13.0 1. 15

1

10

100

0 10 20 30 40

Val

ue

s o

f N

c, N

q, N

γ

φ - Degrees

Nq

Nγ

Nc



The factors for rectangular footing are an interpolation of the square and the strip footing and

are shown on Equation 1.16

qc DNLBcNLBBNq )/3.01()/2.01(5.0 1. 16

It showed be noted that the values of the bearing capacity factors are very sensitive to the

values of shear strength parameters c and υ. Due consideration should therefore be given to

the degree of accuracy of these values. In general the following observations have been made

a) In cohesive soils the contribution of cohesion c to the bearing capacity dominates

b) The depth factor dominates for cohesionless soils

c) The base factor is usually neglected for values of B less than 4 meters

d) A footing at the surface has no bearing capacity if Nγ is neglected

e) The equations are applicable to uniform soils and in the case of stratified soils an

engineering judgment is always required.

Skempton’s values of Nc

Skempton (1951) showed that for a cohesive soil (υ =0) the value of Nc increases with the

value of foundation depth D. He suggested that the values of Nc applicable to circular, square

and strip foundations are given in Figure 1.6. The value of the rectangular footings of

dimensions BxL (where B<L) is the value of a square footing multiplied by (0.84+0.16B/L).

Figure 1.6 Skempton’s values of Nc for a φ =0 soil

4

5

6

7

8

9

10

0 1 2 3 4 5

Nc

D/B

Nc (Strip)

Nc (Circular or

Square)



Eccentric and Inclined loading

Eccentric and inclined loadings have an effect of reducing the foundation bearing capacity.

In the case of a foundation with a vertical load such that the eccentricities are eb and el

(Figure 1.7 ) the effective foundation dimensions are shown as B’ and L’ The resulting load

is distributed over the effective foundation dimensions. The values of B’ and L’ are given in

Equations 1.17 and 1.18

BeBB 2' 1. 17

LeLL 2' 1. 18

In the case of inclined load (Figure 1.8) on a width B and inclination the effective

foundation width is B-2e. In addition the bearing capacity factors are multiplied by the

inclination factors shown on Equations 1.19 and 1.20

2)90/1( o

qc ii 1. 19

2)/1( i 1. 20

1) The base of a long retaining wall is 3m wide and is 1m below the ground in front

of the retaining wall. The water table is well below the base level. The vertical

and horizontal components of the base are 282kN/m and 102kN/m respectively.

The eccentricity of the base reaction is 0.36m. The appropriate shear strength

parameters are c’= 0 and υ’ = 35o.

The unit weight of soil is 18kN/m3.

Determine the factor of safety against shear failure



Solution

For υ’ = 35o,

Nγ = 41 and Nq = 33

The angle of the inclination to the vertical α = tan -1

(102/282) = 20o hence the inclination

factors according to Meyerhof are

The ultimate bearing capacity is given by

The factor of safety

1 m

102kN/m 282kN/

m

282kN/m

1.5m .36m 1.14m



An alternative approach in the case of inclined loads is to use the empirical formula shown on

Equation 1.21

Figure 1. 7 Effective dimensions for pads subjected to eccentric loads

1// ahHavV PPPP 1. 21

Where Pv is the vertical component of the inclined load and PH is the horizontal

component of the inclined load. Pva is the allowable vertical load and PHa is the horizontal

load (a fraction of the available passive resistance).

L

B eB

eL

L’

B’

Y

X



Figure 1.8. Foundation with inclined load

1.2.3 The net foundation pressure

The actual pressure on the soil due to the weight of the structure is called the total foundation

pressure q. The net foundation pressure qnet is the increase in the pressure at the foundation

level. This is the total foundation pressure less the effective weight of the soil permanently

removed during excavation and is given in Equation 1.22

Dqqnett 1. 22

For a strip footing the net foundation pressure is shown on equation 1.23

)1(5.0 NqDcNBNq cnett 1. 23

The safe net bearing pressure (qsafe) is the net bearing pressure factored by an appropriate

factor of safety as shown on Equation 1.24

FOSqq nettsafe / 1. 24

It is usual to use conservative factors of F usually between 3 and 5. Due to uncertainties in

the determination of the strength parameters

and determination of the of the service load,

for comparison the following factors of safety (Table 1.2) are used in other geotechnical

works

e

α P Pv

PH

B



Table 1. 2 Typical factor of safety values for geotechnical works

Failure mode Type of works FOS

Shear Earthworks 1.2-1.6

Shear Retaining walls 1.5-2.0

Shear Sheet piles 1.2-1.5

Seepage Uplift 1.5-2.5

Shear Bearing Capacity 3-5

Effect of ground water

Water table below the foundation level

If the water table is at a depth not less than B below the foundation level the expression for

the net ultimate bearing capacity is given in Equation 1.23 above. However the when the

water table rises to a depth less than B below the foundation level Equation 1.25 is

applicable.

BNNqDCNq subcnet )1( 1. 25

For cohesive soils the value of υ is small and the term ɣsubBNɣ is of little account.

Consequently the bearing capacity is not affected by the ground water variation below the

foundation level. For sandy soils the term CNc is zero and the term 0.5γsubBNγ is about half

0.5γBNγ. The effect of the groundwater is significant.

Water table above the foundation level

For this case the net ultimate bearing capacity is given by Equation 1.26. It is seen both

cohesive and cohesionless soils are affected by the water table rising above the foundation

levels

BNNqpCNq subocnet )1(' 1. 26

Where p’o is the effective overburden above the foundation level.

1.2.4 Allowable bearing pressure

In design, the settlement due to the safe net bearing pressure is computed. If the resulting

settlement is not acceptable then the pressure used in the determination of the settlement is

reduced. At the point when the settlement is acceptable then the pressure obtained is the

allowable bearing capacity of the soil.

In design the ultimate loads are obtained from structural analysis. The ultimate load

is converted into the service load. The gross load is the structural load above the ground floor

plus the overburden. The net load at the foundation level is the load at the ground floor in

addition to the weight of the foundation less any soil which has been replaced. For practical



considerations it is therefore not necessary to consider the weight of the foundation below the

ground level (Figure 1.9)

Figure 1. 9 Net load applied at the foundation level

1.2.5 Field methods for the determination of bearing capacity of soils

Plate bearing test

The test is particularly suited for the design of foundations or footings where it is considered

that the mass characteristics would differ from the laboratory tests and where the precise

values of settlement are required. The plate load test covers the determination of vertical

deformation and strength characteristics of soil insitu. From the data recorded the allowable

bearing capacity of the soil is estimated.

In the test an excavation is made to the expected foundation level. The plate usually

300 to 750 mm square should be rigid and flat. It is loaded by means of kentledge. The

kentledge can be any form of dead load including water, concrete blocks etc or tension piles.

The loading procedure can be either constant rate of loading or incremental loading

procedure as described below:-

Constant rate of penetration

This test is best suited to undrained conditions. In the test the load is applied in a controlled

manner to enable a continuous and uniform rate of penetration. The load is continued until a

penetration of 15% of the plate width is achieved. The ultimate load is considered to be the

load corresponding to the 15% of the plate width penetration.

Incremental load test

This test is best suited to drained conditions. In the test the load an estimate of the maximum

load is calculated. Five equally spaced increments are then selected. The load at each

increment is recorded together with the corresponding settlement. A load is maintained until

the penetration has ceased or when the primary consolidation is complete. The ultimate load

is considered to be the load corresponding to the 15% of the plate width penetration as in the

case of the constant rate of penetration test.

Plate bearing capacity test results

P Gross load =P + overburden

Pnet = P + foundation load – replaced

soil

= P



The plate bearing test results are best reported in graphical way as shown on Figure 1.10.

The weak soft clays and loose sands will reach the ultimate bearing capacity in the region of

100-200 kN/m2 while the stiffer clays and dense sands and gravels will continue increasing

in the bearing pressure with increasing settlement.

Figure 1.10 Typical plate loading test results

Estimation of allowable bearing pressure from plate bearing test results

The test is reliable only if the stratum being tested is reasonably uniform over the significant

depth of the full scale foundation. A weak stratum below the significant depth of the plate

but within significant depth of the foundation would have no influence over the plate test

results but will have a significant effect over the performance of the foundation (Figure 1.11).

-15

-12.5

-10

-7.5

-5

-2.5

0

0 200 400 600 800 1000

Set

tlem

ent

(mm

)

Bearing Pressure (kN/m2)

Stiff clay, dense sand or gravel

Soft clay or loose sand



Figure 1.11 Influence of weak stratum

Settlement of the stratum increases with increasing loaded area and the main

problem is in the extrapolation of the test results to full scale scenario. Ideally the plate test

should be carried out using plates of different sizes and at different depths. However, this is

usually not economical.

Notwithstanding this shortcomings the following procedure was been proposed by

Terzaghi and Peck (1948) and can be used as a guide to use of plate bearing test results. The

settlement of a square footing kept at a constant pressure increases as the footing size

increases. The relationship is shown on Equation 1.27 relates the settlement of the test plate

of 300 mm square and that of a square foundation of width B.

2

1 ))3.0/(2(* BBSS 1. 27

Where S1 = settlement of the loaded area under a 305mm plate for a given pressure

intensity p

S= the settlement of a square foundation of width B in metres under pressure p

In order to use the plate bearing results the maximum allowable settlement is determined. A

value of 25mm is generally accepted as an allowable settlement. S is then equated to 25 and

a numerical value of B is inserted in the formula to enable the determination of the S1. From

the relationship of p and s1 the value of p corresponding to the calculated value of S1 is the

allowable bearing pressure subject to any adjustments certain to the ground water conditions.

Standard penetration test

The test covers the determination of the resistance of soils particularly sand and loose to

medium loose gravels at the base of a borehole to the penetration of a split barrel sampler

when dynamically driven in a standard manner. In addition to the determination of resistance

the split sampler is used to obtain disturbed samples for determination of remolded properties

b B

1.5B

1.5b

Weak stratum



namely particle size analysis and Atterberg limits when the sample has some degree of

plasticity. When used in gravels the sampler is replaced with a 60o cone which does not

sample the soil.

Figure 1.12 shows the main features of standard penetration test equipment. The

drive shoe and the sampler consist of 51 mm external diameter and 35 mm internal diameter.

It is 450-600 mm long. This is connected to a drive assembly at the bottom of the boring

rods. A pick and release mechanism which ensures a free fall of a hammer weigh 65

kilograms through of 760 mm + or – 20 mm is used to drive the sampler or the cone in the

case of the gravelly strata

Figure 1.12 Standard penetration equipment.

The procedure requires that the borehole is cleaned carefully to ensure that disturbed soil at

the bottom of the borehole is removed. When boring below the ground water table it is

prudent to maintain the water in the borehole at the same level or higher than the general

ground water. A hydraulic balance is needed to avoid the risk of boiling of the strata at

bottom occasioned by a high hydraulic gradient.

The sampler and the hammer are lowered to the bottom of the borehole. If after

touching the bottom the sampler penetration exceeds 450 mm on its own weight and that of

the hammer, the SPT value also known as N value is recorded as zero. Otherwise after the

initial penetration on own weight the test is driven in two stages known as seating drive and

test drive

The seating drive is the initial 150mm penetration or 25 blows whichever is reached

first. The test drive is the further penetration of 300mm or 50 blows which ever is reached

first. The number of blows for the 300 mm penetration is recorded as the SPT value ‘N’. If

300 mm penetration can not be reached in 50 blows the test drive is terminated. In this case a

hard stratum has been encountered and further driving results in damage of the split sampler.

It is usual to record the blows for every 75 mm penetration. If the test drive is terminated the

penetration corresponding to 25 and 50 blows is recorded.



Interpretation of Standard Penetration Test Results

The pore water pressure generated by the hammer during testing affects the value of N.

When the test is carried out below the water table in fine sand or fine silt the resistance

increases as a result increased pore water pressure which does not dissipate immediately. If

the measured N if greater than 15 a correction as shown on Equation 1.28 is performed.

)15(2/115 NTrueN 1. 28

The relative density of a soil affects the N values. Terzaghi and Peck (1948) evolved a

qualitative relationship between the relative density and the standard penetration N values.

Gibbs and Hortz put values of relative density. Table 1.3 shows the two relationships

Table 1. 3 Relationship of N values and the relative density of sands

N value Terzaghi and Peck Gibbs and Hortz

0-4 Very loose 0-15%

4-10 Loose 15-35%

10-30 Medium 35-65%

30-50 Dense 65-85%

50+ Very Dense 85-100%

The effective stress at the level of the test also affects the penetration of the SPT split barrel

sampler. This effect can be related to the effective overburden at the level of the testing.

Craig (1986) has summarized the correction of the overburden into Equation 1.29.

NCN N' 1. 29

Where N’=the corrected value of SPT

N=the measured value of the SPT or the true N in the case of the saturated loose

sands and silts

CN=Overburden factor

The relationship of CN and effective overburden is shown on Figure 1.12



Figure 1.12 Estimation of N’ from measured values of N (Craig 1986)

Standard penetration resistance increases with increasing particle size, increasing over-

consolidation ratio and increasing angle of internal friction of the soil. A correlation between

shear strength parameter and N, and effective overburden is shown on Figure 1. It provides

rough estimate of value of and should not be used for very shallow foundations.

Figure 1.13 Correlation between shear strength parameter φ and N and effective

overburden

0

100

200

300

400

500

0 0.5 1 1.5 2

Eff

ecti

ve

Ov

erb

urd

en (

kN

/m2

)

Correction factor CN

0

10

20

30

40

50

0 50 100 150 200 250

SPT

- N

Effective overburden (kN/m2)

φ=25

φ=30

φ=35

φ=40

φ=45

φ=50



Estimation of allowable bearing pressure from standard penetration tests

In 1948 Terzaghi and Peck presented a chart as shown on Figure 1.14 for the estimation of

allowable bearing capacity while limiting the settlement to 25mm and differential settlement

to 75% of the maximum settlement. The procedure involves determination of the average

value of N’ from all the boreholes at the foundation level. The allowable bearing capacity for

the widest foundation is determined and then applied to all the foundations. Terzaghi and

Peck based his chart on foundations on unsaturated soils when the water table is at lower than

1.0B below the foundation. Thus when the water table is at 1.0B the reduction of the

allowable bearing capacity is zero. The reduction increases linearly as the water rises. When

the water table is at the ground level the reduction is 50%. Thus the provisional value of

allowable bearing capacity obtained from Figure 1.14 should be reduced by the factor Cw

shown on Equation 1.30

)/(5.05.0 BDDC ww 1. 30

Where Dw= depth of the water table below the ground level and D

D =the depth of the foundation

B = the width of the foundation

Figure 1. 14 Relationship between N and allowable bearing pressure



Static cone penetration test

The test apparatus consists of a 60o cone as shown on Figure 1. The cone is subjected to

continuous penetration in the soil the rate of 15-20 mm per second. The tip has electrical

sensors for continuous recording of the resistance and penetration as shown on Figure 1. On

the more sophisticated cones the friction along the cone can be measured. In addition the

water pore pressure can also be measured. At every penetration the resistance is measured as

load/cone area and is plotted against penetration

Figure 1.15 Static cone penetration apparatus

Figure 1. 16 Static cone penetration test results

-15

-12.5

-10

-7.5

-5

-2.5

0

0 200 400 600 800 1000

Pen

etra

tio

n (

mm

)

Resistance = load/end area = qc (kN/m2)



From the data the value of Nγ used in Terzaghi Equation for the Ultimate Bearing Capacity

can be estimated from Equation 1.31 the value of internal angle of friction can be obtained

from Figure 1.5 which then enables the determination Nq and the ultimate bearing capacity.

Other empirical values of qa can be obtained from equations 1.32 through 1.33

80/cqN 1. 31

30/ca qq for B< 1.2m 1. 32

2)/)3.0((*50/ BBqq ca for B> 1.2m 1. 33

Allowable bearing capacity on rock stratum

The bearing capacity of rock is the highest that an engineer can expect to get. In some cases

the intact rock has unconfined compressive strength larger than the strength of the concrete

which goes to the making of the foundation. In this case it is the structural design of the

materials rather than the strength of the rock control the foundation design.

For ordinary structures when site investigation is performed by boring, bedrock need

be proved to a depth of three meters to discount the possibility of isolated boulders (Craig,

1987). When un-weathered rock has been reached in foundation construction, the allowable

bearing pressure is based on the inherent strength or the parent rock. The influence of joints,

discontinuities and shear zones is to reduce the allowable bearing capacity. The rock quality

designation (RQD) defined as the ratio of the total length of core of full diameter and length

greater than 100mm or greater to the length of the core run measures the extent of defects and

has been used in the determination of the allowable bearing pressure as shown on Table 1.

Table 1. Allowable bearing capacity RQD

RQD Allowable bearing capacity (kN/m2)

100 29,300

90 19,500

75 11,700

50 6,800

25 2,900

0 1,000

Source Peck et al, 1973

Bowles (1982) stated that the settlement is more often the concern than the bearing capacity.

Consequently most effort should be taken in the determination of modulus E and Poisson’s

ratio η so that an estimate of the settlement can be made. Alternatively he suggested that one

should use a large factor of safety on the unconfined compression strength of the intact

fragments obtained from the borings. The factor of safety should depend on the RQD and

typically range between 6and 10.



Tomlinson and Boorman (1986) reported the presumed bearing capacity must not

exceed half of the unconfined compression strength of the intact rock fragments. Ibi (1986)

reported presumed allowable bearing capacity values of various rocks varying from 12,500

kN/m2 for igneous and limestone rocks to as low as 150 kN/m

2 for weak un-cemented

mudstones.

Rock strength designations based on the unconfined compressive strengths have

been suggested by BS 5930 (Ibi (1986) and the Canadian Geotechnical Society (Franklin and

Dussealt, 1989) are shown on Tables 1.4 and 1.5 respectively.

Table 1. 4 Rock strength designation by BS 5930

Classification Very

Weak

Weak Mod

Weak

Mod

Strong

Strong Very

strong

Extremely

strong

UCS (kN/m2

x103)

Under 2 1.25 to

6

5 to

20

12.5 to

60

50 to

200

100 to

200

Over 200

Source – Tomlinson and Boorman (1986)

Table 1. 5 Rock strength designation by Canadian Geotechnical Association

Classification Extremely

Weak

Very

Weak

Weak Medium

Strong

Very

strong

Extremely

strong

UCS (kN/m2

x103)

Under

2

2 to

6

6 to

20

20 to

60

100 to

200

Over 200

Source: Franklin and Dussealt (1989)

1.2.6 Presumed bearing capacity of soils and rocks

It is common to use presumed bearing capacity of soils and rocks. The values used have been

derived after many years of testing and performance monitoring of existing structures. These

values are usually conservative do not consider the overburden above the foundation level.

They can be used as preliminary values for the very large structures where an accurate

bearing capacity at the foundation level is needed. In the case of smaller structures these

valued can be considered as final. Table 1.6 shows the presumed bearing capacity of soils as

suggested by BS8004 (1986), while Table 1.7 shows the presumed bearing capacity values

used in Kenya. It is to be noted that difficult soils such as expansive soils, loose sands and

silts and made up ground should be investigated all the time.



Table 1. 6 Presumed allowable bearing vales (BS 8004: 1986)

Category Types of soils and rocks Value

( kN/m2)

Remarks

Rocks Strong igneous and gneissic rocks in sound

Strong limestone and strong sandstones

Schists and slates

Strong shales, mudstones and siltstones

10000

4000

3000

The foundations

should be taken to

un-weathered rock

Non

cohesive

soils

Dense gravel, or dense sand and gravel

Medium dense gravel or medium dense sand

and gravel

Loose gravel or loose sand and gravel

Compact sand

Medium dense sand

Loose sand

>600

<200-600

<200

>300

100-300

<100

The foundation

width to be not

less than 1m and

water level not

less than below

the foundation

level

Cohesive

soils

Very stiff boulder clay and hard clays

Stiff clays

Firm clays

Soft clays and silts

300-600

150-300

75-150

<75

Soils susceptible

to long-term

consolidation and

settlement

Very soft clays and expansive clays and silts Not applicable

Peat, organic soils, made up ground and fill areas Not applicable

Table 1. 7 Presumed allowable bearing values in Kenya

Soil and rock Value (kN/m2)

Red coffee soil (Red silty clay) 80-120

Medium dense sand 100-300

Loose gravel (Murram) 100-150

Dense gravel 200-400

Compact gravel and weathered rock 350-600

Un-weathered rock >600

Expansive soils, loose sands and silts Not Applicable

1.3 Proportioning of shallow foundations

1.3.1 Contact pressure distribution

This is the distribution of the pressure below the base of the foundation and the ground. The

pattern of the distribution varies according to the stiffness of the foundation. The stiffness

may be described as yielding (elastic), rigid or flexible



Yielding foundation

The stiffness of such foundation is zero. Here the contact pressure distribution has the same

variation as that of the load. Because of its zero stiffness there will be no moments induced

in the footing. Such a condition exists in fresh concrete before it sets. It has no practical

significance.

Rigid foundations

Contrary to the yielding foundation the rigid foundation has infinity rigidity. They are so

rigid that they do not deflect. Most of the foundations considered in practice are rigid

foundations. The analysis is simple and leads to economical design of the footings.

Flexible foundations

The stiffness of such foundations lies between rigid and the yielding foundations. The

foundations in this category deflect to a certain degree depending on the magnitude of their

stiffness. The analysis of such foundations is complicated but leads to an economical design.

However this is not usually done in practice and is not considered in these notes.

1.3.1 Proportioning the foundations

The proportioning of the foundations is usually the final step in the design of a structure. The

type of foundation, sizes and the level of the foundation depend on the result of the site

investigation. Usually partial factors would have been used in the design of the columns.

However unfactored loads would be used in the proportioning of the foundations. The

factored loads are however required in the determination of the foundation depths and design

of the foundation in accordance with BS 8110 (1997). The general procedure for the design

of the foundations follows the following steps

a) Evaluate the allowable bearing pressure in a site investigation exercise

b) Examine the existing and future levels around the structure and take into account the

ground bearing strata and the ground water level to determine the final depth of the

foundation

c) Calculate the loads and the moments if any on the individual footings with partial

safety factors on the structural loads.

d) Recalculate the loads and the moments on the individual columns and the walls

without partial factors of safety. In many cases it is sufficiently accurate to divide the

factored loads and moments with 1.45.

e) Calculate the plan area of the foundation using unfactored loads

The plan area of the foundations is determined assuming that all the forces are

transmitted to the soils without exceeding the allowable bearing pressure. The distribution of

the pressure is assumed to be planar. In no case should the extreme pressure be less than

zero. All parts of the foundation in contact with the soil should be included in the assessment

of the contact pressure. Subsequently the designer carries out the structural design of the

foundations. Typical foundations are now discussed



Strip and rectangular footings

A strip footing is significantly greater in length than in width. This type of foundation is used

to support walls and closely spaced columns. When and individual column is supported by a

footing then this foundation is referred to as a pad footing. When two or more columns are

supported by one footing, this is referred to as a combined footing.

Axially loaded strip and rectangular foundations

The contact pressure of these foundations is considered as uniform when loaded axially. The

pressure under the foundations should not exceed the allowable bearing pressure of the

supporting soil. Figure 1.17 shows the pressure distribution of such foundations.

Figure 1.17 Pressure distribution below individual and strip foundations under axial

load

Eccentrically loaded rectangular foundations

When foundations are subjected to axial and moments at their foundations the soil pressure

resultant does not coincide with the centroid of the footing. The resulting pressure is a

combination of the compression and the moment stresses. While the columns can in almost

all cases resist the moments it is doubtful that the spread footing can sustain an applied

column moment. The base usually will rotate and induce more moment at the far end of the

column.

In conventional analysis the contact pressure distribution under eccentrically loaded

rectangular foundations (Figure 1.) are derived from the common flexural formula. The

general formula for the estimation of the pressure when there is eccentricity in the y and x

axis is given in Equation 1.34.

yIMxIMAP xxyyyx **),( 1. 34

Where

σ(x,y) = contact pressure at any given point (x, y)

P = the vertical load

x,y = coordinate of the point at which the contact pressure is calculated

My and Mx = the moment about y and x axis respectfully

a) Pad foundation b) Strip foundation

d) Pressure distribution c) Combined foundation



Ix and Iy = moment of inertia of the footing area about the x and y axis respectively

=L*B3/12 and BL

3/12 respectively.

Figure 1. 18 Rectangular foundation eccentrically loaded in two axis

When Equation 1.34 results in negative values in some areas, this means that the foundation

soil is taking tension. It is then necessary to change the dimensions to have only compression

pressure at the base. This is difficult and requires trial and error approach for solution of

maximum and minimum pressures. It is prudent to place the foundation such that that there

is only eccentricity in one axis direction as explained below.

Eccentrically loaded rectangular foundations in one axis

In design it is common to determine the magnitude of the contact pressure at the edges.

Equation 1.34 reduces to equation 1.35 shown below and Figure 1.19 shows the pressure

distribution.

)61( LeBLPq 1. 35

Figure 1.19 Soil pressures below footing

Mx y

L

B

x

My

ey

M

ex

P Mx and My

y

ex

ex

e

L

L

M M

P P

L L L

P

a) e<l/6 b) e=l/6 c) e>l/6

e



When the eccentricity inside mid-third of the base (Figure 1.19a,e<l/6) the computed

minimum pressure is positive soil pressure and the computed maximum pressure should not

exceed the allowable bearing pressure. At e=l/6 Figure 1.19b the minimum soil pressure q=0

and the footing is fully effective in bearing. This limit of eccentricity means that as long as

the eccentricity is less than l/6 also described as falling within the mid-third of the foundation

the entire footing is effective. When the eccentricity is large (Figure 1.19c) and e>l/6 the

computed minimum pressure is negative soil pressure. This is an indication of a tensile stress

between the soil and footing. This in not feasible and the soil pressure has to be evaluated

neglecting any soil tension. The eccentricity is said to be outside mid-third.

For eccentricity outside middle third with respect one axis the maximum soil pressure

redistributes itself since the base cannot take negative pressure. The distribution of pressure

is triangular and is shown on Figure 1.20. The equations applicable in this case can be

derived as follows:-

Figure 1. 20 Eccentrically loaded rectangular out of middle third

eLL

23'

and )'(2

BLq

P

Solving the two equations to obtain the maximum soil pressure q, Equation 1. is obtained

)2/(3*2

elBP

q

1.36

Rectangular combined footings

M

L

L’/3

B

e=M/P

P

L’



It may not be possible to place columns at the centre of spread footings if they are near the

property line, near mechanical equipment or irregularly spaced columns. Columns located

off center will result in a non uniform soil pressure. In order to avoid the non uniform soil

pressure, an alternative is to enlarge the footing and place one or more of the columns in the

same footing to enable the center of gravity of the columns loads to coincide with the center

of the footing (Figure 1. . The assumption here is that the footing is rigid. The column loads

are taken as point loads and distributed into the footing. The footings are statically

determinate for any number of columns. The column loads are known and the resulting

pressure is shown in equation 1.37

APq / 1. 37

Figure 1. 21 Combined rectangular footing

Trapezoidal shaped footings

A trapezoidal shaped footing is required when a combined rectangular footing will not result

in uniform pressure. This is usually so when the space between the combined footings is

constricted as shown on Figure 1.22.

Figure 1. 22 Trapezoidal footing

P2 P1

a b

variable S

X’

L



From Figure 1.22 the position of the centre of area of the footing is x’. The centre of the area

is to coincide with the center of gravity of the loads from the two or more columns being

supported by the trapezoidal footing. The position of the base cannot be extended beyond the

length dimension L. L is therefore a known dimension. The value of the area of the

foundation is obtained from the allowable bearing pressure and the total column loads (

aqPA / ). . The area of the base is shown in Equation 1.38 and the position of the centre

of the area is shown in Equation 1.39. The solution to the two equations leads to unique

values of a and b representing the dimensions of the trapezoidal footing.

Lba

A2

1. 38

babaL

x

2

*3

1 1. 39

From Equation 1.39 and Figure 1.22 it can be seen that the solution for a=0 is a triangular

footing and for a=b it is a rectangle. The solution for a trapezoid footing exists only for

231 L

xL

Strap or cantilever footings

A strap footing is designed to connect an eccentrically loaded column to an interior column

as shown on Figure 1.23. The strap is used to transmit the moment caused by eccentricity to

the interior column footing so that a uniform soil pressure applied to both footings. The strap

serves the same purpose as the interior portion of combined footing and is used in lieu of

combined rectangular or trapezoidal footing. Equations 1.40 through 1.43 are used to

proportion the footing dimensions. The value of eccentricity e is chosen arbitrary by the

designer. Unique solution of the strap footing is not always possible

SPSR 11

1*

111 SS

PR 1. 40

1212 RPPR 1. 41

xeL 2/1 1. 42

aqLBR ** 111 and aqLBR ** 222 1. 43



Figure 1. 23 Typical strap footing

Three basic considerations for strap footing design are:-

a) The strap must be rigid (Istrap/Ifooting>2. This rigidity is necessary to avoid rotation of

the exterior footing.

b) The footing should be proportioned to approximately the same soil pressures and

avoidance of large differential settlements

c) The strap should be out of contact with the soil so that there are no soil reactions and

is weightless

A strap footing is to be considered only as a last option when other options would not work.

The extra labor involved in the forming of the deep beam and accompanying costs make it

only an attractive alternative when other options have been exhausted.

Raft foundations

A raft foundation is a large concrete slab used as a foundation of a several columns in several

lines. It may encompass the entire foundation area or only a portion. Raft foundations are

generally used to support storage tanks, several pieces of industrial equipment or high rise

buildings. Figure 1.24 shows some typical raft foundations

A raft foundation is used where the supporting soil has a low bearing capacity.

Traditionally the raft is adopted when pad and structural wall foundations cover over half the

area enclosed by the columns and the structural walls. However this should be evaluated on a

case by case basis since the raft foundations end up with negative moments and top and

bottom reinforcement. This arrangement could end up being more expensive than closely

spaced pads which require only bottom reinforcement.

R2

x e

R1

P1 P2 S

L1/2 L2

S1



(a) Flat slab; (b) Thickened under columns or beam slab (c) Basement walls as part of the raft or

cellular construction

Figure 1. 24 Common types of raft foundations

The advantages of the raft foundations over the other foundations include:-

a) The effect of combining the column bases is increase in the bearing capacity of the

foundation. This is because the bearing capacity increases with the breadth of the

base.

b) The raft foundations bridge over the weak spots

c) They reduce settlement and are particularly suitable for structures sensitive to

settlement.

Raft foundations are usually designed as infinitely rigid in comparison to the supporting soil.

This assumption simplifies the pressure under the raft to a linearly distributed contact

pressure. The centroid of the contact pressure coincides with the line of action of the

resultant force of all the loads acting on the raft. Figure 1.25 shows the pressure distribution

and the resultant of the vertical loads.

(a

)

(b

)

(c

)



Figure 1. 25 Linear pressure distribution below a rigid raft

A raft foundation is considered as rigid if the column spacing is less than 1.75/λ. λ is given by

Equation 1.44

4/1

**4*

IEbK

c

s 1. 44

Where Ks = coefficient of sub-grade reaction

B = width of strip of the raft between centers of adjacent bays

Ec = modulus of elasticity of concrete

I = the moment of inertia of the strip of concrete

λ. = characteristic coefficient

Bowles (1982) suggests that the coefficient of subgrade reaction be estimated from Equation

1.45.

as qFK **40 1. 45

Where F = the factor of safety applied to the ultimate bearing capacity

qa = the allowable bearing capacity

Equation 1.44 is applicable when the column loads do not vary in magnitude by more than

20%. The column loads should also be uniformly spaced. The design of the raft follows the

following basic steps

a) Compute the maximum column and wall loads

b) Determine the line of action of the resultant of all the loads

c) Determine the contact pressure distribution using Equation 1.46. Figure 1.26 shows

the arrangement of the columns and the eccentricities with respect to x and y axis.

σmax σmin

Resultant of column and wall loads

Resultant of soil pressure



y

x

x

yyx I

xePI

yeP

AP ****

),(

1. 46

Where ∑P=total loads on the raft

A = Total area of the raft

x, y =Coordinates of any point on the x and y axis passing through the centroid of

the raft

Ix and I y = moment of inertia of the area of the raft with respect to the x and y axis

respectively

ex and ey = the eccentricities of the resultant force in the x and y direction

It is conventional to obtain the pressures at the four corners and then interpolate in between to

enable the determination of moments and shears for the structural design of the raft

Figure 1. 26 Raft foundation plan showing column loads

1.3.2 General consideration in the selection of the foundation depth

Once the geometry of the foundation of the foundation has been found, it is necessary to

determine an appropriate depth of the foundation. The following are general considerations

which the designers should take into consideration.

a) Usually the foundation should be placed below the depth with minimum moisture

variation over the years. This eliminates the shrinkage and collapse effects of the

foundation soil. In this country a depth of between 1.0 and 1.5 metres is usually

sufficient.

L

B x

y

ex

ey

P1

P7 P9

P2

P4

P8

P6

P3

P5

∑P

P2 ex

ey Mx

My



b) The foundation should be placed below top soil and below depths with roots of tress.

The roots are potential water paths which weaken the foundations.

c) The foundations should be sited with due consideration to existing nearby structures.

The exaction of the foundation in the vicinity of the existing structures could lead to

loss of lateral support of the neighboring structures.

d) Special attention should be taken to foundations supported on expansive soils and

those on loose sandy silts which are likely to be saturated during the lifetime of the

structure.

e) For water structures viz: - river bridges it is necessary to take extra care to ensure that

scouring of the foundation vicinity does not impair the safety of the foundation. It is

usual to use gabions in areas where scouring is likely to erode the foundations such as

downstream of box culverts and around abutments and pier foundations

f) It is preferable to place foundations at one level throughout. None the less if it is not

practical to have the foundations at one level, the change of level should be at one

plane. Sloping foundation levels should be completely avoided even if they are on

rock. There is a risk of the foundation sliding.

1.3.3 Foundations for common buildings

This section deals with foundations for ordinary common buildings. These are single and

double storied buildings with structural walls as the main form of support. The spans should

generally not be bigger than six metres. The buildings are generally on good bearing soils.

The bearing soils include red coffee soils, gravelly soils and firm sandy, gravelly clays. The

footing for these common buildings is shown on Figure 1.27. The 600 mm width is a

practical width which allows masons to maneuver into the trench.

Figure 1. 27 Typical strip footing for an ordinary building

100mm slab with BRC no 65 at the top face

Damp proof membrane

100-200 mm thick hardcore

600mm wide x 200mm deep

mass concrete foundation

200-150 mm thick

masonry wall

200-150 DPC

A minimum of 1000 mm

depth of foundations

150 mm minimum drop

dropasountonsd



The following are the general considerations in the usage of the standard footing.

a) No reinforcement is needed for strips where the load can be distributed through 45o.

b) The foundations should be excavated and the last 150mm excavation be finalized

when the concreting can be done without further delay. This minimizes the softening

of the foundation

c) The mass concrete is in mass concrete usually by volume batching to achieve grade

15 concrete. A ratio of 1:3:6 for cement sand and ballast respectively is generally

sufficient.

d) Reinforced concrete foundations are done for areas with concentrated loads. These are

usually column supports. Grade 25 concrete is the lowest class of concrete allowed in

the new BS 8110, but grade 20 of concrete can be considered.

1.4 Foundations on difficult soils

1.4.1 Foundations on expansive clays

Introduction

The problems associated with expansive soils arise as a result of alternate heaving and

shrinkage of the clays. These soils are typically black or grey and are referred to as black

cotton soils in this country. The cycle of expansion and shrinkage is a result of ability of the

clays to take in water and retain it in its clay structure. The water absorption leads to

expansion of the clay and causes strains in the foundation and the structures supported

thereupon. The strains eventually cause the cracks to appear on the walls. The result is

structural safety and aesthetics of the buildings are compromised

The clay minerals include montmorillonite, illite and kaolinite as discussed in FCE

311. The montmorillonite clay mineral is particularly prone to heaving and shrinkage. Soil

having more than 20% of montmorillonite are particularly prone to swelling problems

In addition to visual identification the expansive soils can be identified by assessing

the swell potential of the soils. This is done by conducting an odometer test which measures

the free swell and the swell pressure attained in an odometer when a sample held in an

odometer ring is kept at the same volume as swelling is induced by allowing the sample to

take in water. Some of the Nairobi black cotton soils have been found to have a swell

pressure of up to 350 kN/m2.

Chen ( ) has related swell potential to plasticity index as shown

on Table 1.8. The following methods can be applied to mitigate damage control

a) Moisture control

b) Soil stabilization

c) Structural measures



Table 1.8 Relationship of swelling potential and plasticity

Swelling Potential Plasticity index (PI)

Low 0-15

Medium 10-35

High 20-55

Very High Over 55

Source (Chen, )

Moisture control

The main course of heave and shrinkage is the fluctuations of moisture under and around the

structures in question. Depending on the topographical, geological and weather conditions

the natural ground water fluctuates during the year. This seasonal fluctuation decreases with

depth. In some areas the depth to the fluation zone is as low as 1.5 meters. In other areas it

will be deeper going down to over three meters. In addition to the ground water fluctuation

the surface water from rains or bust pipes seeps into the foundations and course moisture

migration.

A satisfactory solution to the problem would to devise an economical way of

stabilizing the soil moisture under and around the structure. It does not matter whether the

moisture is maintained high or low in so far as it can be maintained throughout the year. An

effective procedure of achieving this is to provide a water tight apron of approximately one

metre round the building. A subsurface drain one metre round the building is provided with

augur holes provided at every 2 meters. The holes are filled with sand and interconnected at

the top. In effect the augur drain is and the impervious apron ensures that the moisture at the

foundation area remains the same. Figure 1. 28 shows such an arrangement of the drains for

ensures that the moisture content of the foundations remain the same

The subsurface drain is used to intercept the gravity flow, or; perched water of free

water to lower ground. It also arrests capillary moisture water movement. The subsurface

drain should be lend to a positive outlet. In general the ground surface around the building

should be graded so that surface water will flow away from the building foundations all h the

time.



a) Location of sand drain around a building

b) Sand drain and apron detail

Figure 1. 28 Typical sand drain treatment of a building

Soil stabilization

Soil stabilization consists of one of the following operations

(a) Pre-wetting or flooding the in-situ soil to achieve swelling prior to construction.

(b) Compaction control

(c) Soil replacement

(d) Chemical stabilization

Pre-wetting or flooding the in-situ soil to achieve swelling prior to construction involves the

flooding of the site under consideration prior to construction. The soil would heave and the

potential danger of cracking is eliminated. Pre-wetting has been used with success when the

Compacted granular material at

high water content

Positive drain to outfall

away from the building

Building

Masonry

walling

Ground floor with double

mesh A142

2 meter wide water

tight apron

Original ground

level

Coarse sand drains

at 2 metre intervals

Expansive soil



active zones are not large. It is very difficult to saturate high plasticity clays. There is danger

that expansion of the clays could continue after the construction has taken place. This

procedure should be considered for stabilizing pavement or canal linings. In only rare cases

should the method be considered for use below ground floor slabs. Its application below

building foundations is risky and questionable.

Compaction control has been used in pavement construction. Expansive clays expand very

little when compacted at low densities and high moisture contents. But will expand

considerably when compacted to high densities at low moisture contents. The approach is to

compact swelling clays at moisture contents slightly above their natural moisture content for

good result. In this method it is not necessary to introduce large amounts of water into the

soil. Dry compaction of expansive soils was done along the Lodwar-Kakuma road.

Soil replacement is the simplest an easiest solution for slabs and footings founded on

expansive soils. The expansive foundation soils are replaced with non-heaving materials.

The method requires the selection of the replacement material and the depth to replacement.

In Nairobi the depth of the expansive black cotton soils is in the region of 1.0 to 1.5 metres.

In this case it has been found desirable to remove the entire expansive soil below buildings

and replace with suitable granular material. When the expansive soil is deeper building slabs

can be constructed above the compacted soil covering the expansive soil but the foundation

of main structure needs further consideration.

This method is particularly useful for the construction of highway pavement in a site

completely overlaid with expansive soils where the alternative to reroute the road is not

viable. In this case it the lower expansive soils are overlaid with the compacted replaced

material to a depth of 1.5 metres.

Chemical stabilization is the process of mixing additives like cement and lime to expansive

soil to alter its chemical structure and in the process retard its potential expansiveness. Lime

reduces the plasticity of the soil and hence its swelling potential. The amounts used range

from two to eight percent by weight. Cement on the other hand reduces the liquid limit,

plasticity and potential volume change. Stabilization has been used mainly in highway and

airport construction.

Structural measures include several methods have been reported in literature such methods

include

(a) Floating foundation

(b) Reinforcement of brick walls

(c) Foundation on piles

Floating foundation concept is a providing a stiffened foundation. This is essentially a slab

on ground foundation with the main supporting beams resting on non-cohesive non heaving

material. The slabs are designed fixed on the beams that assuming a heave pressure of 20



kN/m2. This magnitude is small considering that the swell pressure of the expansive soils

commonly found in Kenya has been estimated at between 300 and 500 kN/m2. Results of

such an approach have been mixed where they have been tried. This method needs further

research.

Reinforcement of brick walls have been tried in South Africa. In this method reinforcement is

placed in brick walls. The reinforcement is placed where cracking usually takes place. This

is typically above and below openings. The structure is made also semi flexible by providing

joints in the brickwork so that when heave takes place the building will conform to the new

ground shape and consequently reduce the bending moment induced in the walls. The joints

are typically 1.5cm.

Foundation on piles is a very successful procedure which ignores the heave by placing the

footing to a sufficient depth (Figure ). The depth of the pile should leave an expansion zone

between the ground and the building to allow the soil to swell without causing detrimental

effect to the building. One way of installing the piles is to provide a pile with bell at the

bottom. The bell or under reamed section should be well below the active zone. The bell is

installed with special equipment and anchors the pile into the ground. The pile can be

installed in an oversize shaft which is subsequently filled with straw saw dust as filler to

eliminate uplifting of the pile by heaving soil. Alternatively the pile could be a straight and

the effect of the uplift calculated using Equation 1.47 The friction below the active zone is

utilized in the calculation of the bearing capacity of the pile.

1. 47

Where = the total uplift

D = the diameter of the pile

h = the depth of the pile in the active zone

u = the swelling pressure

f = the coefficient of friction between the pile and the soil

f may be taken as 0.15 while the swelling ;pressure varies between 250 and 500 kN/m2



Figure 1. 29 Pile systems for expansive soils

1.4.2 Foundations on loose sands

Foundations on loose sands are particularly difficult due to the likelihood of collapse in the

event of large storms. The storms result in the realignment of the sand particles and

consequent settlement due to repacking of the sand support. This has resulted in large cracks

in buildings which have been placed on this type of foundation soils. The foundation soils

subsequently loose there bearing capacity and the result is settlement of the foundations. The

superstructure has to absorb the settlement usually with resultant cracks of walls and

structural elements.

A real case story is one of the Garissa teachers college whose buildings were placed

on sand strata. The area is generally dry but when the rain comes, it usually very heavy and

comes in large storms. The performance of the three building types of structures adopted at

Garissa teachers college forms a case study whose findings are used to suggest a construction

procedure for foundations and masonry superstructures on loose sands.

The main teaching bungalow consisted of buildings constructed with a ground beam

which was framed with columns and a concrete roof slab. The masonry was thus reinforced

at the corners with columns and subsequently bound at he top by a ring beam and at the

bottom with a ground beam. These types of buildings were found to have performed well

several years after construction. This type of construction produced a satisfactory type of

constructed and when the buildings were inspected ten years after construction the structural

frames and the infill masonry walls were performing well.

The second type of buildings consisted of three and four and three storied flats. As in

the case of the previous buildings these types of buildings were found to have performed well

ten years after construction

h

h1

Straight pile

Sta

ble

zon

e

Up

lift

S

kin

fric

tio

n

Under ream

pile

Act

ive

zon

e

Act

ive

zon

e

Beam Beam

Sta

ble

zon

e

Up

lift

S

kin

fric

tio

n



The third type of the buildings was the staff residential bungalows. These were

constructed with a ground beam and masonry walls. The roof of the buildings was a concrete

slab. However as the rains came and went in there stormy characteristics the residential

houses developed cracks in the walls. The cracks were particularly severe in the external

walls and after about 10 years of service and needed attention (Plate 1.1

Based on the satisfactory behavior of the framed structures it was found prudent to

introduce columns at the masonry wall corners in a repair scheme. Plate … It is therefore

recommended for foundations on loose sands the masonry should be reinforced with columns

at the corners. In addition the foundations should be kept as far as is possible free from

percolating water. In this way the in the event of settlement the frame will be able to absolve

the stressed attributable to additional settlement and reduce the severity of the cracks.

Plate 1.1 Cracks in the walls occasioned by settlement of the foundation

Plate 1.2 Introduction of columns to stiffen the walls



1.5 Tutorial examples on chapter one

1) A footing 2.25 m square is located at a depth of 1.5m. The strength parameters are c’= 0

and υ’ = 38o.

Determine the ultimate bearing capacity

a) If the water is is well below the foundation level.

b) If the water table is at the surface.

Given that the unit weight of sand above the water is 18 kN/m3. The saturated

unit weight of soil is 20kN/m3.

Ans – A 2,408kN/m3 B, 1,365 kN/m

3

2) A strip footing is to be designed to carry a load of 800kN/m rum at a depth of

0.7m in gravelly sand. The appropriate shear strength parameters are c’= 0 and υ’

= 40o. determine the width of a footing if the a factor of safety of 3 is specified

assuming that the water level may rise to the foundation level Above the water

table the weight of the gravelly sand is 17 kN/m3. The saturated unit weight of

strata is 20kN/m3.

Ans – 1.55m

3) A footing 2m square is located at a depth of 4 m in stiff clay of saturated unit

weight 21kN/m3. The undrained strength of the clay at a depth of 4m is cu=

120kN/m3 and υ’ = 0. For a factor of safety of 3 with respect to shear failure,

what load can be carried by the footing

Ans – 1680kN

4) You are responsible for the design of a combined footing to support two columns

as shown in the figure below. The vertical dead loads on column A and B are 500

and 1400KN respectively. The design requires that the resultant of the column

loads acts through the centroid of the footing. In addition the dead loads, columns

A and B also can carry vertical live loads of up to 800 and 1200 KN respectively.

The live loads vary with time, and thus may be present some days and absent

other days. In addition the live load on each column is independent of that on the

other column. Check that the design meets all eccentricity requirements if the

worst possible combination of live loads is imposed



5) A column is carrying a load of 1200kN. The column is located 300mm form the

boundary of wall. Calculate the pressure distribution if the column is founded on a

square base of 1500mm x1500mm. is the foundation safe if the allowable bearing

pressure is estimated at 300kN/m2

6) An internal column is carrying a load of 2400kN. It is located 3000mm from the

column described in Question 1 Design:-

a. a suitable combined base for the two columns

b. A suitable strap footing for the two columns

7) Your client acquires the next plot and you are not limited by the boundary wall.

Calculate the safe bearing pressure below the columns described in questions 1 and 2.

Assume a detailed site investigation has established the following strength parameters.

C’ = 10kN/m2, υ’ =20

o, γsat = 18 kN/m

2, γb= 16 kN/m

2,

4 Four columns are carrying a tower. If the columns are on a square grid of

2.5mssquare, calculate the pressure at each of the four column positions if a raft

foundation of 3 mmx3m is designed to carry the foundation loads estimated at

4000kN, 5000kN, 6000kN and 7000kN


Chapter two: Deep Foundations

Deep foundation can be categorized into three major types. These include

i. Pile foundations

ii. Drilled piers

iii. Caisson foundations.

The ground and structural conditions which require the use of the two types are discussed

under each of the sections dealing with the two types of the foundations.

2.1 Pile foundations

2.1.1 Introduction

Pile foundations are structural members used to transmit surface loads to lower levels in the

soil mass. They are used when soil beneath the level at an appropriate raft or conventional

footing is too weak or too compressible to provide adequate support to the structure load.

The piles have small cross-section area compared to their lengths. The pile materials

generally include timber, steel or concrete. The transfer is by vertical distribution of load

along the pile surface and at the pile end point.

Piles may be used in the following circumstances

a) To transfer loads to a suitable bearing layer when weak strata is ignored and the load

is transferred to an overlying strong bedrock or compact layer.

b) To transfer load through the shaft friction when compact layer is very deep and would

be impractical to reach it

c) To support structures over water where conventional exaction and construction of the

foundation is not possible or very expensive to achieve.

d) To reduce settlement and in particular differential settlement

e) Based on cost. It might prove economical to drive piles down the strata and then

build on top of the piles instead of having to excavate deep layers and then construct

ordinary foundations

- 46- Deep Foundations


f) In structures which have considerable uplift, horizontal and/or inclined forces. This is

especially true for marine and harbor works.

g) To increase the bearing capacity by vibration and compaction of granular layers of

soil.

h) In soils where deep excavations would result in damage of existing buildings.

Piles can be distinguished by the function they are intended to perform or by the material and

construction procedures used in their construction. The various types of piles by function are

shown on Figure 2.1. The main function of the piles is to take the loads by end bearing or by

friction or by combination of the two. Other functions exist and two which can be sited here

include tension piles and fender piles. The tension piles take lateral forces in place of

traditional retaining walls while fender piles also referred to as dolphin piles are marine

structures principally for taking horizontal loads from vessels in the docking areas. Section

2.2 is presentation of piles by their material and construction procedures.

Figure 2. 1 Types of piles by function

2.1.2 Classification of Piles by materials and construction

Soft soil

Hard

strata

End bearing pile Friction pile

Friction

resistance

Tension resistance

Impact from floating

object

Dolphin or fender pile Tension pile

Soft soil

Soft soil

Firm

strata

Combination



Piles are constructed in a variety of properties of materials, construction methods and

functions. This makes as simple classification difficult. Notwithstanding theses difficulties

they are classified in accordance with the pile materials and method of construction (Figure

2.2). This classification also identifies the pile materials. The principal timber materials are

timber, concrete and steel.

Types of piles

Driven piles Bored piles

Large displacement Small displacement Replacement

Preformed. Solid Cast in place formed Steel sections A void is formed

or hollow tubes by driving closed H Piles by excavation.

closed at the end tubular sections Open ended tubes the void is filled

and left in position and then filling the unless a plug forms with concrete

void as the tube is during driving

withdrawn sides may be

Supported or

unsupported

Solid Hollow

The supporting may effected permanently

Pre-cast concrete or Steel or concrete by casing or

Timber. Formed to tubes closed at the Temporarily by casing or drilling mud

required lengths as bottom. Filled or (Betonite) or

units with mechanical unfilled after driving By soil on a continuous auger

Figure 2.2 Principal Types of piles

a) H and

pipe piles

b) RC

Precast pile

c) Shell

Pile

d) Cast in-situ

tube withdrawn e) Bored pile



2.1.3 Driven piles

To install prefabricated and some form of cast in place piles it is necessary to displace soil by

driving the piles. The piling is commonly done by means of a hammer. The hammer

operates between guides or leads by use of lifting cranes. The leads are carried by the cranes

such that they can drive vertical or raking piles. The piling assembly may be mounted on

base suitable for operation on land or on a floating pontoon in the case of piling in the sea.

The hammers may be free falling operated by a clutch release mechanism.

Alternatively they are powered by diesel or steam. There are several forms of mechanical

devices and equipment in the market used by piling contractors. In order to reduce the

impact stresses on the hammer and the piles it is normal to strike the pile through a hammer

cushion. The elements of cushion vary but are mainly wood packing in a steel cap or dolly.

The various elements in the cushion not only protect the top of the pile but have a significant

influence on the stress waves developed in the pile during the driving. The rating of a

hammer is based on the gross energy per blow. For a drop hammer the rated energy is the

product of the hammer and the height of fall. The efficiency of the hammer is the defined as

the energy delivered at impact divided by the gross rated energy. Energy having been lost in

the dropping of the hammer to pile. For driving piles to great length the hammers have

energies of between of between 50kNm to over 180kNm.

Piles are installed by impact hammers and driven to a resistance measured by

number of blows required in the final stages of piling. For wood piles the energy would be

limited to about 3 to 4 blows per inch when energy of 15kNm is applied by the hammer. If

the pile is to be driven through heaving strata then, it might be necessary to predrill the

borehole where the pile is to be driven. This eliminates undesirable heaving. Additionally if

the pile is to be driven through dense layers of sand and gravel it is possible to loosen the

hard strata by sending a stream of water jet with specially adapted equipment. The various

types of driven piles are now described.

Timber Piles

Timber piles are made of trunks of timber. The timber should be preserved to prevent decay.

Untreated timber embedded below the ground water table has a long life. If the timber is

exposed to alternating wetting and drying it is subject to decay. These types of piles are not

very common.

Steel Piles

Steel piles (Figure 2.2a) are usually in form of H-Piles and pipe piles. H piles are preferred

where high depth is required while the pipe piles are usually filled with concrete after driving.

In the case of H-Piles the flanges and the web are equal thickness in order to

withstand large impact forces. Steel H piles penetrate the ground more readily than other pile

types because of the relatively small cross-section area. They are subsequently used to reach

stronger bearing stratum at great depth. Steel H piles have also relatively large bearing

capacity of between 500 and 2,000 kN per pile depending on the size of the H section. The

pile H sections are usually 250x250 to 350x350 with varying section thickness.



Pipe piles are of the range of 250mm to 750 mm diameter. The wall thickness is

usually over 2.54mm. In the event that the wall thickness is less than 4.54mm the pile has to

driven with a mandrel. When the thickness of wall is over 2.5mm the pipe acts with any

concrete in carrying the load. Pipe piles are usually driven with the lower end closed with a

plate. In some instances conical driving shoes have been attached. The advantage is not

significant.

Steel piles are subjected to corrosion. The corrosion is minimal when the entire pile

is embedded in natural soil. However, the corrosion can significantly increase in the event of

entrapped oxygen. Zones of water table variation are particularly vulnerable. Severe attacks

are encountered on sea structural sections exposed to high and low water tides where the salt

sprays can significantly cause corrosion. The standard practice is to use piles which have a

factory applied epoxy coating. The most vulnerable sections of the piles should be encased in

concrete.

Hard driving and driving through obstructions causes the piles to twist and bend.

They can easily go out of plumb without the piling team recognizing since the depth is at

depth. Deviations from the vertical of below 10% are usually accepted. A penetration of 2 to

2.5mm per blow should be considered as refusal and further driving would generally cause

deterioration.

Pre-cast Concrete Piles

Pre-cast Concrete Piles (Figure 2.2b) are usually cast in a casting yard and transported to the

construction site. Where hard driving is expected the tip of the pile is fitted with a driving

shoe. They are usually of square or octagonal section. The reinforcement is necessary

within the pile to withstand both handling and driving stresses. It is necessary that the exact

length to be installed be determined accurately. If the required length is underestimated, the

extension can be done only with a lot of difficulties. If the length provided proves to be

longer than needed at the site, the piles have to be cut again with a lot of difficulties.

Pre-stressed concrete piles are used and generally have less reinforcement. The pre-

stressing reduces the incidence of tension cracking during handling and driving. The

difficulties related to the pre-cast concrete piles also apply to the pre-stressed concrete piles

Pre-cast concrete piles have relatively large bearing capacity of between 800 and

2,000 kN per pile. The presence of high concentrations of magnesium or sodium sulphate in

the piled environments causes the piles to deteriorate. The deterioration is in the form of rust

in the reinforcement, cracking and spalling. The best practice is dense concrete of high

quality or the use of pre-stressed piles which are not so much susceptible because tension

cracks are minimized.

Driven cast in place piles

Driven cast in place fall in two categories namely case or uncased type. In the cased type

also known as shell the shell type a corrugated steel or pipe which is driven into the ground.

The driving is terminated when the desired length of the pile has been achieved. The

concrete is poured in the shell and left place. In the shell is then left in place. Figure 2.3

shows the schematic installation of a shell type pile.



`

Figure 2.3 Shell type of pile

In the uncased type a steel tube is driven into the ground and tube is withdrawn upon

concreting. Figure 2.4 shows the schematic installation of a typical driven cast in situ pile

where the casing is withdrawn. The pile illustrated is also known as a Franki pile.

Figure 2.4 Installation of a Franki pile

Difficulties encountered in the installation of driven piles

The installation of driven piles has difficulties due to various factors incidental to the

installation procedures and to the ground encountered at the sites. These difficulties are

varied but the main ones include:-

a) Handling of the preformed sections which could lead to damage of the piles before

installation.

(1)

(2) (3) (4)

(1) A gravel pug is compacted at

the lower end of the pile tube

(2) Pile driven to the required

set

(3) Plug broken and a concrete

plug is formed

(4) Core concrete is inserted

(5) Tube is withdrawn as

concrete is placed

(5)

(1)

(2) (3) (4)

(1) RC shells threaded on

mandrel and set in position

(2) Pile driven to the required

set

(3) Mandrel is withdrawn and

top shells above the top of

the pile are removed. A

cage of reinforcement is

introduced

(4) Core concrete is inserted



b) Noise arising from the hammer dropping on to the pile. This can be particularly

undesirable in sites in the busy neighborhoods.

c) Spoiling of the pile in the driving operations include the spoiling of pile heads and or

pile toes. This usually takes place due to overdriving piles when refusal has been

reached. It is usually sufficient to achieve a penetration of 2-2.5 mm per blow in the

last stages of piling.

d) Piles of small cross-section especially H piles driven in boulderly strata could easily

alignment. Vertical piles could end up having bent up shapes and hence lose their

carrying capacity.

2.1.4 Bored piles

Bored piles are also known as cast in place concrete piles (Figures 2.2c-e. The borehole is

effected by various methods using piling equipment. The bore is supported by casing or by

drilling mud (bentonite suspension). At the required depth boring is stopped and the hole is

filled with concrete. If required a cage of reinforcement is placed before concreting is done.

With the use of bored piles larger diameter piles have been installed with corresponding high

bearing capacities. They are constructed in diameters ranging from 300mm to as high as

2400mm. They have been performed to depths of 70 metres and below and can be

constructed vertically or in rakes of up to 1:4. They are thus ideal for many site conditions.

The construction sequence of bored piles depends on the method of construction adopted.

The main construction methods include bored piles with casing support and bored piles with

bentonite support.

Bored piles with casing support

In this type of pile the casing is advanced by a crane and a casing oscillator. The material

below the casing area is excavated and brought up for examination and testing where

necessary. After the depth needed has been achieved the reinforcement cage is inserted

followed by concreting as shown on Figure 2.5

Bored piles with bentonite support

In this type of pile a lead casing is advanced into the soil. The material below the casing area

is excavated and brought up by use of drilling equipment with a bucket which can bail out the

drilled soil. The excavated soil is examined and tested where possible. The drilled hole is

supported by drilling mud After the depth needed has been achieved the reinforcement cage

is inserted followed by concreting as shown on Figure 2.5



a) With casing

b) With betonite support

Figure 2.5 Installation of a bored pile with drilling mud

Difficulties encountered in the installation of bored piles

The difficulties associated with the installation of bored piles are also varied but the main

ones include:-

i. Poor base preparation after the bearing strata has been reached. Loose particles will

have reached the bottom of the bore and will be difficult to detect or remove. The

This installation is suitable in all soils

Install

starter

casing

Advance into the

soil by drilling

and supporting

with bentonite

Insert

reinforcement

cage

Place concrete

with a tremie

pipe and recycle

bentonite

Complete pile

This installation is particularly desirable in gravelly and boulderly conditions

Install casing

using an

oscillator

Advance the

casing and

excavate with grab

Insert

reinforcement

cage

Place concrete with a

tremie pipe as casing

is withdrawn

Complete

pile



base the pile will consequently have a lower bearing capacity than would have been

expected

ii. Poor concreting control where the pile is being cast under artesian conditions. This

usually results from poor shaft control as the concreting continues. The result is

necking of the concrete and/or washout of various sections of the pile. Under ideal

conditions the concreter under tremie conditions should always be placed inside the

wet concrete.

iii. Vibration and movement of the ground in the vicinity of the pile under construction.

It is to be noted that these difficulties are also present in the driven cast in place piles where

the casing is withdrawn as concreting proceeds

2.1.5 Determination of pile load carrying capacity

Determination of load carrying capacity by soil mechanics

Pile design is preceded by extensive site investigation to establish the geotechnical properties

of the soil where the piles will be installed. The parameters obtained in the investigations are

then used in the estimation of the load carrying capacity of the piles. Piles derive their

capacity from base resistance and from side friction. The ultimate load that can be carried by

a pile is then given by Equation 2.1. The terms are explained in Figure 2.6. The accuracy

of the equation depends on the determination of the parameters used in the determination of

Qb and Qs.

Where

= Ultimate Load carrying capacity of the pile

Ultimate Load carrying capacity of the base of the pile

= Ultimate Load carrying capacity of the pile side friction

2. 1

Where

Ab= Area of the pile at the toe of the pile

qf = Ultimate bearing capacity at the toe of the pile

= Surface area of the pile shaft

= Ultimate shearing resistance of the shaft of the pile generally referred to as the

shaft friction

An appropriate factor of safety is applied to the ultimate load. It is prudent to apply different

values for the base and the side friction. This is primarily because the movement needed to

mobilize the friction resistance is much less than the movement needed to mobilize the base

resistance. Initially as the pile is loaded the load is taken by the side friction and as load is

increased the base takes more load. At failure the proportion of load supported by friction



may actually decrease slightly due to plastic flow of the soil near the base of the pile.

Equation 2.2 shows the allowable load when allowing for a factor of safety of 2 and 3 for side

friction and base resistance respectively.

2.2

Figure 2.6 Load distribution of load on a pile

Cohesive soils

Base resistance: The base resistance Qb of piles in cohesive soils is based on the bearing

capacity factor Nc .

2. 3

Where

= bearing capacity factor which is usually taken as 9.0

= undisturbed un-drained shear strength of the soil at the base of the pile

= the cross section area of the pile at the base

In the case of driven piles the clay adjacent to the pile is displaced both laterally and

vertically. Upward movement of the clay results in heave of the ground around the pile and

can cause reduction of the bearing capacity of the pile. The clay in the vicinity of the pile is

completely remolded during driving. Excess pore water pressures are set up during driving.

This pore pressure dissipates in a few months and in any case before significant load is

applied to the pile

In the case of bored pile, the clay area around the pile will be remolded. Additionally

as the water seeps towards the created borehole their softening of the soil in the vicinity of

Qs

Qb



the pile. Water can also be absolved from the wet concrete when it comes in contact with the

clay. The upshot of this is and subsequent reduction of the pile bearing capacity.

Side resistance is based on the friction mobilized on the surface of the pile. Equation 2.4 and

2.5 shows the estimation of the side friction

2. 4

2. 5

Where

= adhesion factor between the pile and the soil

= the average undisturbed shear strength of soil adjoining the pile

= the shaft area which contributes to the friction resistance

Most of the load of a pile installed in a clay soil is derived from the shaft friction and the

problem usually revolves accurate determination of the value of α. For soft clays driving of

piles tend to increase strength around the pile. A value of α equal to 1 can be used. It is

however unlikely that the soil will not in the long run return to its original soft status after

some time. In over-consolidated clays the value varies from 0.3 to 0.6 (Smith and Smith,

1998). A value of 0.45 is usually used for design purposes.

An alternative is approach is to express skin friction in terms of effective stress. The

rationale of this approach is that the area of disturbance during pile installation is relatively

small. The excess pore water pressure induced in the installation process dissipates ahead of

the application of load.

2. 6

Where

Ks = the average coefficient of earth pressure and

= the average effective overburden pressure adjacent to the pile shaft

= the angle of internal friction of the remolded clay. The cohesion intercept of

remolded clay in an drained triaxial test being zero.

Cohesionless soils

Base resistance: The ultimate bearing load carried by a pile depends mainly on the relative

density of the sand in which it is driven. The ultimate bearing capacity at the base of the pile

is given by



Where

= The bearing capacity coefficient.

= The effective overburden pressure at the base of the pile

It is to be noted that the bearing capacity attributable to Nγ usually ignored in pile design as

the value of B is usually small. The values suggested by Berezantzv et al (1961) are often

used and are shown on Figure

Figure 2.7 Bearing capacity factors for use in pile design

Source Berezantzv et al 1961

Side friction: Meyerhof (1959) suggested the average value of friction to be estimated from

Equation 2.6. As can be seen from the Equation the value of fs continues to increase as the

effective overburden increase. However field tests have shown that the maximum value of fs

occurs when the embedded length of the pile is between ten and twenty diameters. In practice

a maximum value of 100 kN/m2 of fs is taken.

2. 7

Where Ks = the average coefficient of earth pressure and

= the average effective overburden pressure adjacent to the pile shaft

= the angle of internal friction between the soil and the pile.

Typical values of and Ks are given on Table 2.1 after Smith and Smith (1998) are shown on

Table 2.1. The ultimate load that can be carried by the pile is therefore given by Equation

2.7.

10

100

25 35 45

Val

ue

of

Nq

φ in Degrees

Nq



Table 2.1 Typical values of and Ks

Pile material Ks

Loose Dense

Steel 20o 0.5 1

Concrete 0.75υ 1.0 2.0

Wood 0.67 υ 1.5 4.0

Source Smith and Smith (1998)

2.8

Equation 2.8 shows the allowable load when allowing for a factor of safety of 2 and 3 for

side friction and base resistance respectively.

2.9

Determination of piling parameters from in-situ tests

The above equations pose difficulties with respect to determination of parameters for a

cohesionless soil which is difficult to sample in the field in undisturbed condition for accurate

determination of Nq which depends on the internal angle of friction. The value of the angle of

internal friction between the soil and the pile remains at best an estimate.

Consequently it has been found preferable to use empirical correlations based on the

results of standard penetration and those of the Dutch cone penetration equipment. Meyerhof

(1976) proposed the values given on Table below.

Table 2.2 Pilling parameters from standard penetration tests

Driven piles

Type of soil qb (kN/m2) fs (kN/m

2)

Sands and gravels

Large diameter -

Average diameter -

Non plastic silts

Large diameter -

Average diameter -

Bored piles

Any types of soils

0.67

Source Smith and Smith (1998)

Where N = the uncorrected blow count at the base of the pile

= the average uncorrected value of the blows over the embedded length of

the pile

D = is the embedded length of the pile in the bearing stratum



B = the width or the diameter of the pile.

An alternative to the use of the Standard Penetration tests is to use the Dutch cone test results.

The cone penetration results can be seen in

Figure 2.8. The ultimate base resistance is taken as average value of Cr over a depth of 4d as

shown on

Figure 2.8. The ultimate skin friction can be obtained from Table 2.3.

Figure 2.8 Typical results from a Dutch Cone Test

Table 2.3 Skin friction (fs) values from Dutch cone test results

Type of pile fs kN/m2

Driven piles in dense sand

Driven piles in loose sand

Driven piles in non plastic silts

Where

is the cone resistance along the embedded length of the pile

The allowable bearing load of the pile as before based on the Dutch Cone Test results is

given by Equation 2.9

2.10

Cr (kN/m2)

Dep

th (

m)

Est

imat

ed d

epth

of

the

pil

e

3d

d



2.1.6 Determination of load carrying capacity dynamic methods

Determination of load carrying capacity dynamic methods is applicable to driven piles. The

basis of derivation of dynamic formula is that a relationship exists between the pile capacity

and the driving behavior during the last stages of driving. The energy from the hammer to the

pile is transformed into useful energy and can be represented by Equation 2.10 in the last

stages of the pile driving

2. 11

Where

M = the mass of the hammer

g = the acceleration of the hammer

h = the drop the hammer

R = the pile capacity

S = the settlement of the hammer as result of the drop h

In practice the above Equation has been modified to take account of several losses which take

place during the driving process. The main losses of energy occur as a result of sound, heat,

friction, quake, losses associated with elastic behavior of the pile and those associated with

the pile head compression. The net energy is equated to the work done in penetrating the

ground by the pile. Figure 2.9 shows the sequence of the pile driving and the

a) Variation of energy upon falling of hammer on to a driven pile

b) Penetration of pile upon falling of hammer on to a driven pile

efW

h efeivWh

Wh

(sso+ses)

(sso + spp) +(sep +ses)

Permanent +Elastic penetration

(sso)

(ses)

(sso + spp =set =s)

(sep +ses )=c)

h



Figure 2.9 Energy and penetration of a pile during driving

The potential energy of the hammer is Wh. Upon contact with the pile the available energy to

drive the pile into the ground is ef.eiv.Wh, where ef is the efficiency upon falling and eiv is

the efficiency upon impact. The penetration of the pile as shown on Figure 2.9b can be

shown to result in permanent ;penetration attributable to the pile and soil spp aand sso. In

addition there will be elastic penetration sep and ses attributable to the pile and soil

respectively. The work done and the pile resistance equation can now be rewritten as shown

on Equation 2.11.

2. 12

Where R = The ultimate load capacity of the pile

= the overall efficiency factor

Equation 2.10 is known as Hiley formula. In the field the final stages of the pile are

monitored and recorded as can be seen on

Figure 2.10. It is usual to drive the piles to a minimum set of 2.5mm. Harder driving only

goes to damage the toe of the pile and could reduce the pile capacity in the process. Pile

driving formulas should be used in the piles driven in sand and gravel and in any case should

be calibrated with a load test.

Figure 2.10 Pile driving trace of the final stages

set = s1 set = s2

Elastic comp = c2

set = s3 Elastic comp = c1

Elastic comp = c3



2.1.6 Determination of load carrying capacity pile testing

The load test is the most reliable of all the methods used in the determination of load carrying

capacity of a pile. In this method a full scale test is carried out on a working pile. Essentially

the pile is loaded and a plot of load versus settlement is recorded. From the plot the

allowable load is computed by one of the many formulas available from literature. Full scale

piles are then constructed to the same specification as the test pile

The test is conducted by loading the pile with kentledge load or by use of tension piles

(Figure 2.11). In some piling contracts the working piles cannot be used as tension piles for

testing purposes. This is primarily because in the cause of piling test the tension piles are

lifted slightly. This could lead to weakening of the working piles.

a) Load resisted by kentledge

b) Load resisted by tension piles

Figure 2.11 Methods of testing piles in the field

If the test pile is a purely test pile ahead of the main installation of the pile the maximum load

to be applied is equal to two and half times the estimated safe carrying capacity of the pile.

It is usual to load the pile to 1.5 times the design allowable pile load when a working pile is

tested for ascertaining the integrity of the piles installed.

Maintained load test

The load is applied by maintaining the load in a series of increments. The increments are

usually equal to 20 to 25percent of the design working load of the pile. The subsequent

increments are carried out when the settlement has reduced to less than 0.25mm per hour.

The load is subsequently withdrawn in the same stages as the loading to trace the unloading

curve.

Constant rate of penetration

Kentledge

Existing ground level Support

Jack

Test pile

Jack

Test pile

Tension pile Tension pile

Existing ground level

Kenteledge



In this method the load is applied by a constant rate of penetration by a jack in order to

maintain a constant penetration rate (Figure 2.11b). it is usual to maintain penetration rates

of 1.5mm per minute and 0.75mm per minute in the case of sands and clays respectively.

Interpretation of test results

The results are plotted on a load settlement curve as shown on Figure 2.12. In the two

procedures ultimate pile load is taken as the load which achieves a settlement equal to 10

percent the diameter of the pile as is seen in test pile a Figure 2.11b. (BS 8004). The ultimate

pile load could also be reached when the shear failure of the pile soil interface or the pile toe

occurs (Figure 2.12b). The allowable pile load is obtained by dividing the ultimate load by

an appropriate factor of safety. The factor of safety usually ranges from 1.3 to 2.0

a) Maintained load test results

b) Constant rate penetration test results

Figure 2.12 Pile test load results

The above failure criterion is applicable to normal size piles. In the case of large diameter

piles on rock the ultimate load depends on the capacity of the concrete. This depends on the

stress in the concrete.

2.1.7 Negative skin friction

Negative skin friction is a phenomenon or which occurs in piles when a force develops

between the pile and the adjoining soil in a direction which increases the load on the pile and

or the pile groups. This phenomenon develops when a compressible layer of clay, silt, or

Load

Time

Set

tlem

ent

Load

Set

tlem

ent

a

b

Load

Ultimate

load (a)

Ultimate

load (b)

Penetration

Penetration =

0.1 pile diameter



mud etc settles on account of consolidation which may be initiated by ground water lowering

or increase in overburden pressure.

As clay layer settles, piles are dragged into the soil by the consolidating soil and the

overburden soil. The direction of the friction is reversed increases the load on the pile. The

friction generated on the perimeter of the pile due to this dragging is carried by the column

instead of assisting in carrying he pile load. The effect is to reduce the carrying capacity of

the pile. This is the phenomenon known as negative skin friction

Figure 2.13. The negative skin friction may be estimated from Equation 212 for

single piles and Equation2. For group piles

Figure 2.13 Negative skin friction

2.13

For cohesive soils fs is can be approximated to . while for cohesionless soils fs is

equal to . Where the value of fs is estimated from triaxial testing for cohesive

soils the fs can be taken as 0.5Cu

Where

= the ultimate force generated by the negative friction

= the shearing resistance of the soil

= length embedded above the bottom of the compressible layer

= the pile diameter

= the coefficient of earth pressure at rest

= angle of shearing resistance in terms of effective stress

= average effective overburden pressure

Fill

Compressible clay

Len

gth

of

sett

ling s

oil

=l

l-fi

ll

l-cl

ay



2.1.8 Pile groups

In practice piles are designed and constructed to work in groups. In construction of a group a

pile cap is cat on top of the piles. The cap is usually in contact with the soil on top of the

piles. The bearing capacity of the group is an arithmetic sum of the piles and that of the cap.

Banerjee (1975) showed that the pile cap could support up to 60% of the applied load. If the

cap is clear of the ground surface piles in the group are referred to as free standing piles.

Bearing capacity of groups

Except for the large diameter piles of over 700mm diameter the piles are usually designed in

groups of three or more piles under a column. The minimum under a foundation wall would

be two per typical cross-section. Typical arrangement of the piles is given on Figure 2.14. In

general the ultimate load capacity of the pile group is not the sum of the loads of the piles in

the group. The ration of the ultimate load for the group to the sum of the loads carried by

individual piles is the efficiency factor of the group.

Figure 2.14 Typical arrangement of pile groups

For piles in sand, the group action is complicated by dilatancy and densification

characteristics of the sand. When the spacing of the piles is less than eight times the pile

diameter, group action takes place (Department of Navy, Naval Facilities Engineering

Command, 1982). In dense sand the effect of driving piles is to loosen the sand and hence

the angle of internal friction of the sand in the vicinity of the piles. This results in overall

reduction of the pile bearing capacity. The group efficiency factor is less than one. In loose

sand the effect of driving piles is to increase the density of the sand. The bearing capacity of

the loose sand will therefore be increased. In this case the efficiency factor is more than one.

An efficiency factor of 1.2 is often used. In the case of bored piles in sand the resulting

loosening of sand in the boring operation results in efficiency factors less than 2/3. The

difficulties in the quantification of the design parameters of either loosened or densified sand

strata in piling operations remains a real problem for engineers (Mwea, 1984). Nonetheless

3 – Pile 4 – Pile 5 – Pile

12 – Pile



experimental evidence has it that the piles at the centre of a group in sand carry more load

than the piles on the periphery.

For piles in clay the effect of the pile group is to reduce the bearing capacity of the

pile group. This is because the effect of placing piles in a group is to have one large block

taking friction on the sides and base resistance over the block base. The spacing of piles in

clay is of the order of two times the pile diameter to four times the diameter. The efficiency

of the groups range from 0.6 to unity as the pile spacing increases from two diameters to four

diameters. The ultimate load in the case of a pile group is given by Equation 2.13. In the

case where the pile cap rests on the ground the ultimate load should be taken as the less of the

block capacity or the sum of the individual piles on the group.

2. 14

Where = The width of the group

= Length of the base of the group

= Depth of the group

= Bearing capacity factor of the clay

= The average undrained strength of the undisturbed clay

Whitker (1957) in a series of model tests showed that block failure as a group in clays occurs

when the spacing of the piles is not more than 1.5d apart. General practice is however to

space the piles at between 2 and 3d. In such cases the efficiency of the group is

approximately 0.7.

Settlement of groups

The settlement o a group of piles can be estimated by assuming that the entire load acts at a

depth as an equivalent raft. In clays the raft is assumed to be located at a depth of 2/3 D

where D is the depth of the pile group. The load is at spread of 1:4 from the underside of the

pile cap to allow for friction transfer. After the assumed depth of the raft the load is

distributed at a spread of 1:2 (Error! Reference source not found.a). Immediate settlement

and consolidation settlement can then be estimated for the layers of soil below 2/3D by

application of normal methods.

For groups in sand the equivalent raft is at a depth of 2/3Db from depth 2/3D. The

spread from the perimeter of the piles is 1:4 followed by a spread of 1:2 Error! Reference

source not found.b). The settlement of the underlying sand stratum is then gotten from

application of standard penetration data and or the cone penetration resistance

D 2/3D 1:4

1:2 1:4 2/3Db Db

2/3D



Figure 2.15 Equivalent raft concept for piles

2.2 Drilled piers and Caisson Foundations

2.2.1 Drilled piers

The term drilled pier foundations is used in a number of situations which to refer to deep

foundations which method of construction is fundamentally different from that of piles. A

large shaft performed in soil and then filled with concrete may be termed as a drilled pier.

ACI (1972) refers to all shafts where a person may enter and work as a drilled piers. In this

definition all shafts larger than 750mm diameter can be referred to as drilled piers. Figure

*** shows typical piers used in practice. In general drilled piers are used where the soil has

a low bearing capacity and it is necessary large loads to firmer stratum and the following

conditions preclude the use of smaller piles.

i. Pile vibrations are not acceptable.

ii. Pile members are too small for the loads.

iii. A large bearing end is needed for higher load capacity

2.2.2 Caisson Foundations

The term caisson is also used to refer to box type structures consisting of many cells built in,

concrete or steel or combination of both. They are built wholly or partly at higher ground and

sunk to final position. They are used to transmit large loads through water and soil to firm

strata. They are used in large bridges, shore protection structures. They are generally used

under the following conditions.

i. The soil contains large boulders which would otherwise obstruct the penetration of

piles and or construction of cast in place piles.

Straight pier Underreamed pier Pier socketed Into

Rock

1:2

Clay stratum Sand stratum

Position of equivalent raft Position of equivalent raft



ii. A massive substructure is needed to extend below the river bend to provide resistance

against floating objects and scour.

iii. Foundation is subjected to very large lateral forces.

Caissons may be divided into three categories

i. Open caissons

ii. Pneumatic caissons

iii. Box caissons or floating caissons

Open caissons

An open caisson essentially consists of a box open at the top and bottom ( Figure 2.16). the

soil is removed from the caisson by grabbing, dredging from inside the caisson. The sinking

of the caisson proceeds by the caissons self weight assisted by cutting edges of the walls.

When the desired level has been reached concrete is poured under onto the base of the

caisson by tremie pipe. In some cases the caisson has been pumped out. But in most of the

cases the caisson has been left in place. The bearing capacity of the soil below is usually

determined by normal bearing equations.

The concrete seal at the bottom is placed as a plug at the bottom of the caisson but

later serves as a permanent base of the caisson. Its thickness can be obtained from the

equations below

For circular caissons

√

For rectangular caissons

√

Where

= thickness of the seal

σo = contact pressure or hydrostatic pressure

R = radius of the caisson in the case of circular caisson

fc = the allowable concrete stress in tension (0.1 to 0.2cube strength)

b= width or the short side of the caisson in the case of a rectangular caisson

l= length or the long side of the caisson in the case of a rectangular caisson

β = coefficient which depends on the l/b ratio

Water level



Figure 2.16 Open Caissons

Pneumatic caissons

Pneumatic caissons provide an airtight enclosure (Figure 2.17). In effect water is prevented

from getting into the enclosure and the workers can excavate and pour concrete under dry

conditions. The reliability of the quality in this case is better in so the mechanical ventilation

is carried out to the strictest of the specifications. Pneumatic caissons are costly and should

be considered only with the following conditions in mind:

i. Premium pay because of associated health hazards

ii. Overall safety requirements are high

iii. Much of the effort is towards making the work environment suitable for the workers

When the excavation has reached the desired stratum the concrete is sent down to the

working chamber carefully to fill any weak points on the exposed strata. After this initial

filling the area is filled except a small portion of the chamber below the roof of the chamber.

This final portion is filled with grout which also fills any spaces which might have been left

behind during the concreting.

The seal design and estimation of the bearing capacity is the same as that of the open

caissons

Figure 2.17 Pneumatic caissons

Cutting edge

Ground surface

Circular open caisson Box caisson

Compressed air in

working chamber



Box caissons

Open caissons are usually cast on the ground and then towed to the site. They area then

lowered to a prepared ground. They are carefully aligned on place and then made stable by

placement of ballast. The design and construction of box caissons do not bring any new

design requirements. The ground upon which the caisson is being laid needs to have been

exhaustively investigated to ascertain the foundation depth and any likely difficulties likely to

be encountered. After the caisson is in place it may be filled with either sand concrete or

sand. The caisson should be checked against stability as it is floated to the final place of the

intended foundation.

Design of caissons

The caissons will be designed to resist vertical loads including superstructures, own weight

minus buoyancy forces. The lateral forces will typically include forces due to wind,

earthquake, earth and water pressures, and traction from traffic and pressure from current

flow.

The forces acting on a caisson must be estimated as accurately as can be to enable a

safe design. There are many methods adopted by various geotechnical engineers but the for

stability of the caisson the following combination of forces will suffice

i. All forces are resolved into

ii. A single vertical force

iii. Two horizontal forces in the direction across and along the caisson.

It has been found out that analysis of the caisson in a direction transverse to the direction of

the axis is more critical. From Figure ***-* the three equations of static equilibrium are

solved. This are

W = Base reaction + skin friction

Q = Passive pressure created on BF – Passive pressure on DE – Base friction

Q (H+D) = Moment of all the forces

Q Q



QFrom structural analysis H

D

A

F

CE B

W

D1

From geotechnical analysis

γD(Kp-Ka) γD(Kp-Ka)

O

Qmax =Area ABC-Area FEC

Qmax =1/2 γD2 (Kp - Ka)- ½*2* D (Kp - Ka)*D1

Moments about O:

Qmax (H+D)=1/2 γD2 (Kp-Ka)D*1/3- ½*2* D (Kp-Ka)*D*D1*1/3

Therefore D1 and Qmax can be calculated and necessary adjustments of the caisson are

made depending on values of Kp and Ka

From structural

analyses

From geotechnical

analyses

h

D

O



2.4 Examples of Piling Schemes

Sutong bridge in China

Sutong bridge in China (Plate 1), which has a centre span of 1088m, designed in an area of

high winds and likely to be hit with massive earthquakes (Bitener et al, 2007). The

foundation strata presented the designers with particularly difficult task. The soils at the site

consisted of firm to stiff clay extending to 45 metres below the sea bend. This clay strata was

underlain with a medium to very dense coarse sands, silty sands and occasional loam layers

matrix to a to of 250 metres below the sea bed where the basement rock was encountered.

The designed pile groups covered a plan area of 113.8x48.1m. The design consisted

of 2.8 and 2.5 diameter piles. Permanent casings were installed to a depth of 40 metres. The

overall depth of the piles was of the region of 110 metres. The shafts were designed to mainly

be carried by friction since the displacement needed to mobilize the end bearing is two to

three times that needed to mobilize the skin friction The tips of the pile shafts were however

grouted to increase the bearing capacity of the piles. This procedure densifies the soil below

the shaft and any debris left during the drilling operations. The increased the pile capacity

end bearing capacity is of the order of 20%.

Plate 1 of the Sutong Bridge in China (1088 m center span)

The Nyali bridge in Mombasa

This is a pre-stressed concrete bridge founded on seabed which had coral deposits, sand and

clay soils matrix proved to a depth of 100metres below the sea bend. The designers

depended on the skin friction for the centre piers. The design consisted of 2.0metre diameter

shafts drilled down to depth of 50 metres. On plan the piles have a rectangular layout of 3x8

piles per pier.

2.5 Tutorial examples on chapter two

1) A single pile 0.6 m diameter is bored into sand strata six meters thick overlying a clay

stratum of infinite depth. Detailed investigations have established that N value in the

sand zone increases with depth (n=3Z). The undrained cohesion increases with depth

(Cu = 5+4Z). Assuming the adhesion factor α = 0.35, determine



a) An equation for the estimation of pile working load if the pile is to terminate

in the sand zone.

b) An equation for the estimation of the pile working load if the pile is to

terminate in the clay zone.

2) A precast reinforced concrete pile measured 450mm x450mm. The pile was driven to

a depth of 15 metres to a set of 3mm by a drop hammer of 2.5 tones freely through 1.5

metres. The piling arrangement was changed to have a 4.2 tone hammer falling

through 2 metres. Assuming the same resistance with the new hammer, determine the

set achieved if the following information is also available.

2.5 tone hammer 4.2 tone hammer

Overall efficiency factor 0.5 0.35

Elastic compression of pile 4mm 4mm

Elastic compression of soil 4.5mm 5.0mm

3) A pile under test has started showing considerable settlement under load of seventy

tones. The pile diameter is 500mm and a length of 8.5metres in stiff clay. Assuming

below the 8.5metres the clay was soft clay and did not contribute to any resistance

evaluate the magnitude of the unit shear along its skin. (Answer 10.5tones per m2).

4) A 500mm diameter bored pile is to be made in stiff clay to a depth of 20metres. The

un-drained strength of the clay varies with depth as shown in the following table

Depth 4 6 8 142 16 20 24

Cu (kN/m2) 78 86 102 132 157 184 212

Determine the maximum load that may be applied to the pile. The following factors

may be taken.

Adhesion factor α = 0.45

Overall factor of safety = 2

Nc for piles is usually taken as = 9

(Answer 1025kN).


Chapter Three: Introduction to Earth Dams

3.1 Introduction

Advances in geotechnical engineering have enabled design and construction of high dams

impounding large amounts of water. The design and construction follows well documented

procedures gained over the last years from design construction and monitoring of both

successful and unsuccessful projects. The procedures now taken include

i) Thorough pre-design and preconstruction investigation of the dam foundation

conditions and of the construction materials and design of dams.

ii) Application of engineering skills and techniques to design

iii) Carefully planned and controlled construction

iv) Carefully designed and installed instrumentation and monitoring of the completed

dams

The design and construction of a dam is not complete without accomplishing its intended

purpose and has proved it safe over several cycles of the performance. Carefully designed

and constructional dams are in excess of three hundred meters high. Our own Thika dam

which supplies the Nairobi residents with water rises some sixty three meters above its

foundation.

Failures in dams have been occasioned by improper design, inappropriate

construction methods, including preparation of foundations, placement of the dam

embankment layers, without the necessary controls of compaction control and monitoring.

The design and construction should not be stereotyped on existing dams. Rather each dam

should be unique and dependent on the geology of the available materials. As one embarks

on the design of dams it should the course of dams the causes of failure of dams has been

listed by Singh and Prakash (1985) as shown on Table 3.1

- 74- Introduction to dam design


Table 3.1 Causes earth dam failures

Cause of Failure % Occurrence Type of failure

Overtopping 30 Hydrological

Seepage effects (piping, sloughing etc) 25 Geotechnical

Slope slides 15 Geotechnical

Conduit leakage 13 Geotechnical/Structural

Damage to slope protection 5 Maintenance

Miscellaneous 7 General

Unknown causes 5 General

Source: Singh and Prakash (1985)

The design and construction techniques covered in this chapter are applicable to all dams.

However the design and construction of small dams in Kenya is well covered in the manual

prepared by Ministry of Water (1985). Small dams are those whose height does not exceed

15 metres and or its impounded volume does not exceed one million cubic meters (Bureau of

reclamation 1985). The procedures covered in this chapter are inappropriate for the design

and construction of dam materials presenting the followings characteristics

i) Extremely soft, or dispersive or materials with high plasticity

ii) Exceedingly pervious foundations

iii) Exceedingly fractured foundations

These conditions require specialized testing and analysis of the presenting conditions in order

to arrive at an appropriate design

3.2 Selection of type of earth dam

The scope of dams covered is those dams where the major portion of the embankment is

constructed in successful layers compacted in layers. The layers are well bonded into one

another to achieve the necessary requirements of the particular layer. The materials are

borrowed from borrow pits and from the reservoir area of the dam. Earth dams fall into three

categories namely, diaphragm, homogenous and zoned

3.2.1 Diaphragm types

This type of dam is constructed with pervious materials namely sands, gravels and or rock.

An impervious diaphragm is constructed to act as the main barrier to seepage. The

diaphragm is usually made of concrete, or bitumen. Alternatively they are made of thin

compacted earth. In this case the width of the diaphragm at any depth is either less than three

meters or it thickness at any elevation is less than the height above that elevation. Figure ***

shows typical diaphragm type dams



As with all dams the diaphragm dam should be designed and constructed with care

and precision. All internal diaphragms whether made of rigid materials like concrete or even

compacted earth have potential of cracking caused by differential movement of induced

during consolidation of the dam embankment materials, fluctuating water levels or settling

foundations. Internal concrete diaphragms can not be readily inspected. Earth diaphragms

on the surface require protection with filters, protection against erosion and wave action.

These types of diaphragms are unusually protected by rock fill and rock riprap. The earth

diaphragm is also not readily inspected during routine or emergency inspections. The earth

diaphragms are usually protected from internal erosion by filters usually in the form of

geotextiles.

If most of the material in a diaphragm dam is rock, then this type of dam is referred to

as a rock dam discussed below.

3.2.2 Homogenous types

These types of dams are made up of single kind of material save for the slope protection. The

material in this type of dam must be sufficiently impervious to act as the barrier for the

seepage. Because the impervious materials are inevitably clays which are weak in stability

but good as barrier to the seepage the slopes tend to be rather flatfish. The usual slopes on

the upstream side of the dams are 1:3.5 to 4 while for the down slope slopes need slopes of

1:2.5 to 1:3. Figure *** shows a typical homogenous slope with three flow lines. As can be

seen, seepage inevitably appears on the downstream side at a height of about 1/3 of the height

of the dam.

Rock toes and horizontal blankets are usually used to avoid the seepage breaks on the

down slope side of the dam. Riprap protection is also used on the upstream side to arrest

erosion occasioned by the waves on the upstream side of the dam. Drainage and filter layers

are designed to meet filter requirements. Inclined filters in combination with horizontal filters

built with well graded sand and surrounded by geotextiles have become a normal practice.

Because modification of the homogenous dams has led to successful dams the use of

completely homogenous dams is now not allowed. The homogenous dams are preferred

where other materials of contrasting permeability are unavailable. Alternatively they should

be used where impervious material forming the embankment is abundant and available

principally in the dam area and within the vicinity of the dam.

Zoned types

In this dams, a central core is of impervious material is franked by more pervious materials.

The design of these dams requires that the permeability of dam embankment materials

increases from the core to the outside franking shells. The materials enclose support and

protect an internal impervious core. The upstream sections provide stability during rapid

drawdown. The downstream pervious materials act as drainage to control the line of seepage.

It is usual to place a filter material between the impervious material and the downstream

pervious materials.



The impervious inner layers are basically clays typically the red coffee soils. The

pervious layers are sands, gravels, cobbles, boulders and rocks. If a variety of soils are

available the type of dam of choice is the zoned dam (Bureau of reclamation 1985). It has

inherent advantages of stability and reduced seepage across the dam wall.

3.2 Design Principles

The dam should be constructed so that a satisfactory performance at minimum cost is

attained. The maintenance costs should also be factored to ensure a facility with the least

maintenance of the upstream, downstream and the apartment structures and the electro

mechanical structures. An earth dam must be stable during all phases of the construction and

the operation of the reservoir. To accomplish this, following criteria must be met:

i) The embankment, foundation, abutments, and reservoir sides must be stable and

should not develop unacceptable deformations during construction or during the usage

of the structure

ii) Sufficient seepage control must be ensured to ensure that excessive piping, instability;

sloughing, material erosion is under control. Additionally the loss should be such that

it dose not impair the intended usage of the facility by excessive loss of water.

iii) The reservoir sides should be stable under all operating conditions to prevent

landslides into the reservoir. It is to be noted that a landslide into the reservoir could

cause large wave to overtop the dam

iv) The embankment must be provided with adequately sized spillway which allows

design flow floods to pass without overtopping the embankment.

v) Free board allowance should be sufficient to prevent waves from overtopping the

dam.

vi) The dam should be provided with camber which allows settlement of the foundation

and the abutment to take place. This camber is not included in the freeboard

calculations.

vii) The upstream slope must be protected against the wave action while the down slope is

protected against rain erosion and animal grazing

3.3.1 Foundation design

Foundations of dams refer to the dam embankment wall floor and the sides of the

embankment in touch with the original ground of the dam. Foundations are usually not

designed but they require attention to ensure satisfactory performance. The requirement of

the foundation is to be stable under all conditions and to offer sufficient resistance to seepage

to prevent loss of water.

To determine the seepage and stability conditions of foundations the permeability of

the foundations strata in various directions and at various depths need to be determined. In

addition the strength of the strata should be determined by use of appropriated field testing

accompanied by field testing. For small dams however it is normal to use empirical approach



in the treatment of the foundations. Because the foundations of different materials demand

different treatments the foundations are grouped into three different classes. These classes

can be grouped into

i) Rock foundations

ii) Foundations of coarse grained materials (sand and gravel)

iii) Foundations of fine grained materials (silt and clay)

Rock foundations

Ordinarily the rock foundations do not present any bearing capacity problems. Instead it is

the seepage problems which have to be addressed A thorough site investigation should be

undertaken to establish faults and any areas of excessive weathering which could lead to loss

of water. The procedure would be usually to perform in-situ tests to determine the

permeability of the rock structure. This is undertaken together with a site survey of the

fissures of the rock. If excessive erosive leakage , uplift pressures, high water pressures can

occur though rock crevices, fissures, permeable strata, and/or fault planes, consideration

should be made to grout the foundation.

The foundation grout is basically injection of a sealing material under pressure into

underlying formations through grout holes. Grout ordinarily consists of cement water

mixture in the ration of 10:1 in the case of rich mix to 0.8:1 in the case of a lean mix. Some

additives to the cement water mix is usually done to improve the pumping. The most used

additive is betonite

The injected grout eventually fills the cavities and potential avenues of water.

Grouting is a procedure requiring specialized personnel and equipment to effectively carry

out the operations. In general a centerline curtain of grout of holes spaced at three to six

meters is adequate. Where large zones of fracture occur below the dam wall and in the

immediate upstream of the dam a blanket grout on grid is desirable. The depth of the grouting

is usually in the region of three to ten metres. In most cases a blanket grouting of the

foundation directly below the impervious zone is desirable.

Sand gravel foundations

Generally these foundations have sufficient strength to adequately support the loads induced

by the embankment and the reservoir. Nonetheless exploratory and analysis of the strata

must be carried out as a matter of routine. The main problems of these foundations are under

seepage and subsequent forces exerted by this seepage. These undesirable effects should be

analyzed and mitigated in design and construction.

Foundations on looses sands are suspect and should generally be avoided as the sand

has the potential of collapse under load. These type of foundations should be avoided or

specialized advice sought.

The amount of under seepage should be estimated from values of coefficient of

permeability of the strata. The coefficient of permeability of the strata should be determined

by established methods including pump out tests, tests conducted by observation of boreholes

when pumping is performed in a test borehole or pump in tests as described in FCE 311. The



magnitude of the seepage forces should also be determined by analyzing the flow net of the

water flow under the dam. This topic has been covered in FCE 411.

The various methods of treatment of the foundations of sands and gravel should aim

at economical control of the under seepage and the control of the subsequent seepage forces

to prevent the undesirable effect of foundation erosion and piping at the exit of the dam.

Excessive treatment of a detention dam might not be necessary while treatment of foundation

of a water supply dam might be prudent. The various treatment techniques are now presented.

i) Cutoff Trenches

Where possible this is usually the treatment of choice. The cutoff should extend down to

bedrock or to other impervious strata. This treatment ensures no future difficulty will be

experienced in piping and or uplift of the dam. A minimum width of the cutoff trench is

shown on Figure 3.1

Figure 3.1 cutoff trench

ii) Partial cutoff

A cutoff that does not go all the way to the foundation, rather it is designed to a proportion of

the depth to the rock or to impermeable layer.. the reduction in area is not proportional to the

reduction in the flow. Thus the reduction cannot be estimated from the flow equation

The action of the partial cut off is similar to that of an obstruction of in a pipe. The reduction

in flow is not proportional to the reduction in the area of the pipe. Experiments have shown

that a 50% cutoff results in 25% reduction of the seepage while an 80% cutoff results in 50%

reduction in the seepage.

iii) Sheet piling

This is an expensive method of cutting of the seepage through the foundation of an earth

dam. Additionally the seepage continues to pass through the sheet pilling interlocks. It has

been used sometimes in conjunction with the cutoff trenches. The sheet piles cannot be

performed in cobbles and boulders

iv) Slurry trench

h

d

w

Sand gravel

Rock



This is a trench excavated and filled with concrete below the impervious layer. The trench is

kept in position by placing bentonite before concreting to form

v) Grouting

Various materials have been used to develop grouting procedures to improve the stability and

reduce the permeability of pervious foundations under dam walls. These materials include

a) Cement – water

b) Cement – bentonite – water

c) Bentonite

It is to be noted that grouting is usually an expensive process and it should be allowed after

extensive testing and evaluation.

vi) Upstream blankets

These are usually made of same material as the impervious core material. In effect the path

of the underseepage is increased and hence the loss of water is reduced.

vii) Downstream embankment toes

The aim of these blankets

a) To reduce uplift pressures at the exit of the dam

b) To readily permit discharge

c) To prevent piping of the fines

d) To convey the discharge

Achieved by

a) Extending the downstream zones

b) Pervious downstream shells or by use of horizontal blankets

c) By use of toe drains

3.3.2 Embankment Design

Embankment slopes

The design of and earthfill dam embankment needs combination of many parameters. Many

of these parameters are difficult to determine accurately. These parameters include gradation,

composition and corresponding behavior of the soils under different conditions of saturation

and loading. The stress – strain relationships can be very complex. The result of these

difficulties is that the design of earthfill dam embankment depends largely on successfully

designed, constructed and well performing dams.

Notwithstanding these acknowledged difficulties it is now possible to model out complex

conditions of an earthfill dam embankment. The design of any earthfill dam is preceded by

extensive site investigation to determine the strength and permeability characteristics of the

embankment materials. This enables the design of h the slopes to the embankments to be

checked under the follow conditions.



i) Stability during construction and end of construction. In this condition the

embankment has not had the time to have the pore water in the foundations and the

embankments drained. The pore water pressures are highest in the embankment

materials. The strength parameters applicable are the undrained parameters.

ii) Steady seepage conditions. The core of the dams act as the water barrier of the earth

fill dam. However even the tightest of the clay cores will allow some water

penetration. The rate of penetration will depend on the permeability of the core

material an in due time will reach steady seepage conditions where a phreatic surface

will be developed at the highest level in the embankment. The steady seepage

conditions is critical for the downstream slope. Under these conditions the water has

been impounded the seepage has stabilized through the embankment. The flow net

has been established. All the excess pore water pressures have dissipated. The slopes

of the dam are checked using drained parameters of the foundations and the

embankment materials. The downstream slope is in critical condition during the

steady seepage

iii) Rapid drawdown conditions. Under these conditions the stabilizing effects of the

water in the reservoir has been removed on the upstream slope. The rapid drawdown

leaves high pore-water pressure in the embankment. The upstream slope is usually in

its weakest state. The upstream slope of the dam is checked using drained parameters

of the foundations and the embankment materials. It is to be noted that a drawdown

of up to 40 meters per day is considered as rapid.

iv) Stability under severe seismic conditions. The above conditions should be subjected

to acceleration of the embankment occasioned by seismic activities

v) Protection against erosion. The upstream slope is likely to be subjected to erosion

arising out of the wave action and sloughing as the level of the water fluctuates. This

is mitigated by use of appropriate upstream protection by use of stone riprap. The

downstream slope is subjected to erosion a result of the precipitation and made worse

by grazing in some dams in communities in need of pasture. The usual practice is to

fence off the dam area and to plant grass and appropriate trees.

The stability check is usually to ensure that the shear stresses induced in the embankments

are resisted by the mobilized shear strength. The shear stresses are from the externally

applied loads which include reservoir weight and earthquake forces. Additionally internally

generated forces from the self weight of embankment The shear stresses at the slopes being

checked can be shown on Figure **** below the shear stresses to be resisted is shown on

Equation 3.1



3.1

The external and internal forces produce a compressive stress along the sliding surface. This

mobilizes the shearing strength which resists shearing along the surface being checked. The

shearing strength is given by Equation 3.2

3.2

It is to be noted that while the shear strength is reduced by the increase in the pore water

pressure the shear stress remains the same. This shows the need of understanding and taking

care of the changes in the pore water regime. In practice the design involves the checking of

the slope stability and application of a suitable factor of safety

Compaction

Compaction of earthworks is a key activity to ensure that the envisaged strength and water

tightness is achieved. When the compressibility and loading of the embankment are constant

the more saturated the soil is the higher the likely hood of developing high pore-water

pressures. To minimize the development of high pore-water pressures it is compact the

earthworks just dry of optimum. However for low dams it has been found satisfactory to

compact earthworks at MDD and OMC.. At this moisture content the material is able to

conform to the shape of the foundation and the abutments.

3.3 Inspection of existing dams

General appearance

i) Sagging crest

ii) Slope failures

iii) Wet patches

iv) Slope protection

v) Soil erosion – gullies

vi) Loss of riprap

vii) etc

σ τ

σ1

σ3

θ



Spillway

i) You might wish to recalculate the adequacy of the spillway. This topic is

covered separately under hydrology

ii) Check field indicators of adequacy of the spillway – water marks

iii) Blockages

iv) Is the gear control working

v) Structural failures in the concrete

vi) Note any cracks

Gauge house

i) Are the instruments in good working order

ii) Have they been vandalized

Reservoir area

i) Assess the siltation

ii) Assess the conservation measures being undertaken in the neighborhood of the

dam

iii) What is the state of the fence of the reservoir fence for the fenced reservoirs?

AOB

3.4 Examples of earth dams in Kenya

In general the dam axis should be chosen in such a way as the material required for the

embankment is minimal while getting the maximum storage. Usually this is so where the

contours are narrowing downstream of a wide valley. The dam axis should be designed as

straight as possible unless the topographical features dictate otherwise

The height of the embankment should be determined in order to achieve the desired

storage with an increased gross freeboard. The gross freeboard is the height between the

spillway crest and the embankment crest and takes account of the design flood and the wave

height

The crest width should be such that earthmoving equipment can be able to work on

the crest. In many cases a road should cap the embankment. In any case a minimum width

of four meters should be observed.



Table 3.2: Design statistics for Ndakaini dam

Thika district

1 Description

Dam name Ndakaini (Thika)

District Thika Dam type Zoned embankment Designer/Engineer Howard &

Humpreys 1. Catchment area

Catchment area

(km2)

71 Altitude at dam site

(masl) 2000

Mean annual rainfall

(mm) 1500

General soil types Grade IV to VI 3. Embankment

Crest length (m) 420 Crest height (m) 65 Crest width

(m) 11

Bottom width (m) Upstream side slope 3:1 Dow stream side slope 2.5:1 Freeboard

(m) 2

Embankment

volume(Mm3)

2.5

Slope protection Riprap 4 Core

Depth (m) 2 Side slopes 1:1.5u/s ,1:5d/s Core slope protection Filter drains 5. Foundations

Soil type Weathered

rock 6. Reservoir

Fetch (m) 4250 Depth (m) 41 Area (m

2) 2900000

Capacity (Mm3) 70

7. Bellmouth Spillway Crest diameter (m) 15 Shaft diameter (m) 2 capacity (m

3/s) 417

8. Draw off system Height of stand pipe(m) 20 Pipe Diameter (mm) 5500 Height of tower (m) 70

(a): Embankment details

Core1

3.5

12.2

2.25

2.25

2.25

2.751

Downstream shoulder

Filter drain

70 m deep grout curtain

1

1

1

Drainage layers

upstream shoulder

Full storage level 2041 masl

70000

Drainage blanket

11.5

Draw-off tower

culvert intake

outlet

Draw-off pipe

Original Ground level

2015 masl

2030 masl

2045 masl

2025 masl

2005 masl

1985

2000 masl

11000

10000

31

2.51

31

4000

Crest

4000

5000

3000

3000 5

1

masl



(b) bellmouth spillway

Figure 3.2(a): Embankment details of Ndakaini Dam

5.5m pipe

16m diameter belmouth

embankment



Table 3.3: Design statistics for Kwa Tabitha dam, Kitui district

1. Description

Dam name Kwa Tabitha

District Kitui

Dam purpose Domestic water

supply Designer/Engineer NWCPC

1. Catchment area

Altitude at dam site

(masl)

1094


(mm) 720

General soil types Not available

3. Embankment

Crest height (m) 10.5

Crest width

(m)

5

Bottom width (m) 38

Upstream side slope 3:1

Dow stream side slope 2:1

Freeboard

(m)

1.5

Embankment

volume(m3)

12000

Slope protection Hand placed

riprap 4 Core

Width (m) 5

Depth (m) 1

Side slopes 1:2 u/s, 1:2d/s

Core trench volume

(m3)

200

5. Foundations

General soil type Rock

6. Reservoir

Depth (m) 7.1

7. Spillway

Width at sill (m) 15

Depth (m) 2

Excavation

(m3)

25000

8. Draw off system

Height of stand pipe(m) 10

Pipe Diameter (mm) 200

Figure 3.3: Embankment details of Kwa Tabitha Dam, Kitui district Dam

13

12

5000

1000 5000 1000

crest protection murram

Toe drainsand filter

trough

Hand placed riprap300mm

Protective gravel 300mm

Core

Gross freeboard

Normal water level ( 1100 masl)

1500

500

1101 masl

2300

Crest

Grassing

pipeDraw off

10m stand off pipe

Cattle

5000



Table 3.4: Design statistics for Birica dam, Nyeri district

1 Description

Dam name Birica

District Nyeri

Dam type Embankment

Designer/Engineer NWCPC

1. Catchment area

Altitude at dam site

(masl) 2161


(mm)

1500

General soil types Not available

3. Embankment

Crest length (m) 138

Crest height (m) 7

Crest width

(m) 5

Bottom width (m) 44

Upstream side slope 2.5:1

Dow stream side slope 2.5:1

Freeboard

(m)

1.5

Embankment

volume(m3)

22500


riprap

4 Core

Trench Width (m)

Depth (m)

Side slopes

Core trench volume

(m3)

5. Foundations

Soil type Rock

C’

6. Reservoir

Depth (m) 4

Area (m2) 22000

Capacity (m3) 50000

7. Spillway

Width at sill (m) 15

Depth (m) 2

Excavation

(m3)

1800

8. Draw off system

Height of stand pipe(m) 2

Pipe Diameter (mm) 200

Figure 3.4: Embankment details of Birica Dam, Nyeri district Dam

12.5

12.5

Normal water level

Gross freeboard

Original ground level

2169 masl5000

1500

5000

300mm Protective gravel

300mm Hand placed riprap

7000 8000 15000 8000

sand filter

Crest

2m long stand-off pipe



Table 3.5: Design statistics for Kwa Kasenga dam, Machakos district

1. Description

Dam name Kwa Kasenga

Dam District Machakos

Dam type Embankment

Designer/Engineer NWCPC

1. Catchment area

2. Catchment area

(km2)

360000

3. Altitude at dam site

(masl)

4. Mean annual rainfall

(mm) 900

5. General soil types Not available

a. Embankment

Crest length (m) 120

Crest height (m) 7

Crest width

(m) 7

Bottom width (m) 41

Upstream side slope 3:1

Dow stream side slope 2.5:1

Freeboard

(m) 2:1

Embankment

volume(m3)

18000


riprap

4 Core

Trench Width (m) 5

Depth (m) 1

Side slopes 1:2 u/s, 1:2d/s

Core trench volume

(m3)

600

5. Reservoir

Fetch (m)

Depth (m) 4.5

Area (m2) 15000

Capacity (m3) 7018

6. Spillway

Width at sill (m) 12.5

Depth (m) 2

Length (m) Not available

Excavation

(m3)

1800

7. Draw off system Not available

Height of stand pipe(m)

Pipe Diameter (mm)

Figure 3.5: Embankment details of Kwa Kasenga Dam, Machakos district

1 3 1

2.5

4000

1000 5000

500 1500 Gross freeboard

Normal water level

Crest 7000

5000 1000

300mm Hand placed riprap

300mm Protective gravel Homogeneous embankment

7000

sand filter


Chapter Four : Site Investigation

4.1 Introduction

Site investigations are also referred to as soil exploration. It consists of investigating the

condition on which construction is planned. From site investigation it should be possible to

obtain information for the following geotechnical engineering activities

i. Design of new foundations

ii. Modification of existing foundations

iii. Location of materials of construction of roads, runways, etc

iv. Identification of materials needed for the construction of pavement structures for

roads, runways etc

v. Identification of ground to be excavated in the construction of various facilities

including water pipe lines, building foundations, earthworks in cut areas etc

The site investigation should form a part of a coordinated chain of design from inception of

the project through preliminary to the final detailed design of a civil engineering project. It

should indeed continue post construction monitoring of the completed schemes. Because of

the diversity of civil engineering schemes a set of standard procedures is not possible for all

site investigations. The varying civil engineering schemes require a variety of options in

breadth and detail needed for the various schemes. The objectives for which a site

investigation is carried out also differ with various schemes. The main objectives of carrying

out a site investigation are now presented

i) Suitability of site for particular works

In the case of option of site for particular works a detailed site investigation should be able to

enable determination of the most suitable site. Thus it is possible to shift a bridge from one

location which would call for expensive deep foundations to one where ordinary shallow

foundations would be sufficient.

ii) Adequate and economic design

A site investigation leads to safe structures during and after construction. Additionally

sufficient information is obtained for quantifying the excavations needed in the preparation of

the bills of quantities. This should minimizes the possibility of cost overruns due to

unexpected ground conditions being met at construction time.

- 89- Site Investigation


iii) Planning construction

By identifying different materials along the construction paths and their locations a

systematic procedure of carrying out the works is evolved. In the case of road works

materials from the cut areas are analyzed for use in the fill areas. It is then possible to

proceed with construction of the fills and cuts methodically with minimum haulages and

waste of materials.

iv) Prediction in changes in structure

Carefully and well executed site investigations should enable the prediction of the likely

settlement of structures under construction. Equally important is the ability to predict the

effect of excavations on the neighboring structures.

v) Safe structural design of large structures

Heavy modern structures require more detailed site investigations. Today we are seeing

higher buildings, larger bridges and installations sensitive to settlement. Structures and civil

engineering schemes are being put up very quickly. Immediate and consolidated settlement

is taking place when the works are commissioned. Further settlement takes place during the

useful life of the civil engineering installation. Accurate estimation of the settlement regime

is particularly important considering that clients are becoming more and more sensitive to the

performance of structures and the argument that cracks are minor and do not pose any

danger to the structure is no longer good.

4.1.2 Planning a site investigation

Table 4.1 shows a schematic way in which various activities with respect to site investigation

can be performed at various stages of a project. It is clear from the table that site

investigation should not be treated as an afterthought but rather should grow with the project

from conceptual initial design to eventual post construction period.


Table 4.1 Stages of a site investigation

Phase Pre-construction Construction Post Construction

Stage

Conceptual

Initial design Preliminary design Detailed design

Supervision of

construction

Operation &

Maintenance

Main activity

Conceptual

design Design Alternatives Detailed Site Construction control -Performance

Site

investigation

activity

Define Scope of

SI

Desk study of SI –

Review of existing

data Preliminary trial

pits

Detailed investigations

-Boreholes

-Trial pits etc

Laboratory and field

tests

Field observations

– field densities

- field moisture contents -

Monitoring and

checking performance –

- pore water pressures

Settlement

Inclinations

SI Reports

Terms of

reference and

bid documents

i) Preliminary SI

investigation report

ii) Cost estimate of SI

Detailed design report

-SI report As built SI report -

-Maintenance reports

-Performance reports

-Research reports


4.2 Preliminary and detailed stage site investigations

4.2.1 Preliminary stage site investigations

This should lead to information needed for the design of the various alternatives at the

preliminary stage of the study. The activities in this stage can be summarized as follows:

i) A study of any existing site investigation reports for the area or in the neighborhood

should form the basis of this stage of investigations.

ii) A study of geographical a geological maps of the site in the case of large sites.

Topographical characteristics should lead to useful information such faulty areas.

Heavily forested areas are an indication of deep rooted top soils.

iii) A site inspection of the existing buildings and any existing structures. Any signs of

distress which can be related to the settlement of the foundations. Any information

from archives, previous records held by the local authorities.

iv) Inspection of the soil profiles, in cut areas, old used quarries. Structured questions to

local people with regard to the geotechnical information being sought yields

considerable information. Such questions are:

a) What is the depth of the pit latrines in the area?

b) At what depth murram encountered?

c) At what depth was water struck?

v) Aerial survey of the site could give useful information with regard to land formations

and soil profiles.

vi) Seismic refractions could be carried out at this stage of investigations. Usually a

specialist is needed to interpret the results.

vii) Preliminary trial pits

Geophysical methods

Geophysical methods involve sending of seismic or electrical waves through the ground. The

determination of the soil strata is based on the fact that the velocity or the resistance seismic

wave transmission or resistance to electrical flow differs with different rock types and soils.

The method allows the boundaries of the soils to be determined seismic refraction is

described below

Seismic refraction is conducted by having a source of seismic waves (Figure 4.1). The

seismic waves are induced by detonating a small explosive or by striking a metal plate hard.

Waves are subsequently emitted in all directions, through the air, and through the soil in all

directions. Seismic wave transducers called geophones are placed radially from the

epicenter. A circuit connects the geophones and the detonator for accurate determination of

time. A direct wave will reach the geophone first since it is the shortest distance covered.



When there is a dense stratum at depth a refracted wave will travel along the top of the bed

rock. As it travels it leaks energy to the surface which can be picked by the geophone.

Figure 4.1 Seismic refraction – arrangement of equipment

For short distances the direct waves reach the geophones first. For longer distances

the refracted wave reaches first though the distances is longer than t he surface direct

distance. This is so because the speed of the wave in the dense material is higher than that in

the overburden material of less density. The geophone has a mechanism which records the

first wave and ignores the others. This enables a plot of arrival time versus the distance.

The first section of the graph represents the direct wave measurements while the second

section represents the refracted wave measurements (

Figure 4.2). The inverse of these curves are the velocities of the seismic waves. The

general types of the rocks are determined by geophysics from the knowledge of velocity

versus rock type. It is also used in the determination of depth to water table and thicknesses

of multiple strata. The depth D to the bedrock can be estimated from the formula.

Figure 4.2 Time versus distance for seismic waves

√

4.2.2 Detailed stage site investigations

At this stage the aim is to obtain data for use in the final design of the works. The

investigation is carried out by use of trial pits, sounding and boring. The extent of the use of

these methods depends on the type of the project at hand and the geotechnical parameters

being sought.

Tim

e

Distance d

Geophones

Seismic source



The trial pits

The pit and shaft technique supplies the most detailed and reliable data on he existing soil

conditions. Once the trial pit has been dug stratification of the soil should be done usually in

the field. In addition as much information should be recorded. This information includes

i. Depth to ground water table.

ii. Field assessment of the bearing capacity.

iii. Depth of the various strata encountered in the trial pit.

iv. The encountered soils should classified by visual inspection

a. Coarse grained soils should be described with adjectives such as angular,

rounded with traces of fines etc

b. Fine grained soils should be studied to indicate whether they are loamy, of low

plasticity, whether they are sandy clays etc

c. All soils should be described indicating their color and odour if any. Decaying

organic matter if encountered should be mentioned.

v. Obtain undisturbed samples when you can for the different layers of strata

encountered. These samples can then be taken to the laboratory for tests

For large sites the pits should then be surveyed and located in a grid system for incorporation

into the site investigation report.

Sounding tests

These are basically are penetration tests carried out to supplementing trial pits and borings.

The penetration resistance is measured and related to the bearing capacity. They are widely

used in site investigations. They consist of the cone penetrometer already presented in

chapter 1. The other commonly used penetration equipment is the dynamic cone

penetrometer used in the estimation of the California bearing ratio (CBR) of road pavement

layers. This enables the design of the pavement layers to be carried out

Boring methods

When a deep stratum has to be investigated it will usually be necessary to perform boring

operations to ascertain the strata below the ground to be used in the support of the proposed

structures. Several boring methods are available and are summarized as follows

Percussion drilling consists of a derrick, a power unit and a winch carrying a light steel cable

which passes thorough a pulley. The unit can be towed by a vehicle after the assembly is

folded. The assembly drops a chisel on the ground and strata being drilled



Figure 4.3 Schematic presentation of a drilling chisel

The excavation is effected by the drilling chisel. The drilling rods provide the necessary

weight for the penetration the strata. Further weight may be added when need arises. The

winch raises and lowers the chisel and its attachments

Below the water table the loosened soil forms slurry. Above the water table water is

introduced to form the slurry. Periodically the slurry is bailed out by a shell or a bailer to

make progress into the soil. In boreholes which are liable to collapse the borehole must be

cased. In some cases the casings slide on their own weight. On completion of the job, the

casing is jacked out.

Percussion drilling is usually done in diameters of 150mm to 300mm. the borehole

depth investigated by this drilling method can be up to 50 to 60 metres. This method of

drilling can be done on virtually all types of soils including those with boulders and cobbles.

The rig is versatile enough to place mechanical augers and penetrating testing equipments at

appropriate depths.

Power operated augers are usually on vehicles. Downward pressure is applied by pressure

or dead weight. The augurs are 75-300mm diameters. Augers are usually used in self

supporting soils. Casing is usually not needed since the augers have to be removed before

driving. In full flight augers the rod and the helix cover the entire length being investigated.

The augur is then brought up. The soil is ejected by reverse rotation. The likely hood of soil

from different strata being mixed up is very high. In the short flight augur the auger is

advanced into the soil and then raised. The soil is also ejected by reverse rotation.

Rod

Chisel



Figure 4.4 full flight and short flight augurs

The continuous flight augurs are sometimes fitted with a hollow stem which is plugged

during the drilling operations. When samples are needed the plug and the rods are removed

and a sampler is introduced for the recovery of a sample. The sample may be undisturbed

depending on the sampler utilized. The flight augurs are not suitable for use in loose soils

which are likely to collapse as the augur is inserted and removed from the hole.

Hand and portable augers are usually operated by persons by turning the handle of the

augur. The hand augers are typically of 75 – 300mm diameters. The soil is locked in the

auger and frequent removal is needed to ensure that the augur does not get stack in the soil.

Undisturbed samples may be obtained by introduction of small diameter tubes which are

hammered into the strata under investigation. This method is suitable for self supporting

soils. It is not possible to penetrate coarse granular soils.

Figure 4.5 schematic representation of a hand augur

Wash boring is a method of boring where water is pumped through boring rods and released

through narrow holes in the chisel attached at eth lower ends of the boring arrangement

(Figure ****).

Full flight augur Short flight augur



Figure 4.6 schematic representation of a hand augur

In this method the soil is loosened and broken by water jet. This is aided by the up an down

movements of the chisel. An attachment to the rods called a tiller enable the rotation on the

drilling bit. The drilling winch is able to raise and lower the chisel and hence get the

chopping action of the chisel.

This method is suitable for most soils but progress is slow if the particles of coarse

gravel larger particles are present. The accurate identification of the soil types is difficult.

The method cannot be used to recover soil samples for testing. However tube samplers can

be advanced into the borehole for obtaining relatively undisturbed samples.

Rotary drilling is done by use of drilling bits that cuts and grinds the subsoil or rock at the

bottom of the borehole. Water is usually pumped down hollow rods passing under pressure

through to the drilling tools. This cools and lubricates the bits. The fluid also provides

support for the borehole where there is no casing.

Two methods of rotary drilling are available. The first is open drilling where the soils

and rocks are broken within the diameter of the hole. Subsequently the tubes are removed and

tube samplers and testing continues below the borehole. This advances the drilling. The

second method is known as core drilling and involves creation of an annular hole in the

material and intact rock enters the drilling core. This advances the drilling and enables

Water from pump

To sump

Drilling bit

Tiller



samples to be retrieved from the borehole. The sample is then subjected to immediate field

description and taken to the laboratory for various tests. Typical core diameters range from

41mm to 165mm. The method is fast, but in large gravelly soils the speed is slowed by

rotation of the bit without advancement into the ground.

4.2.3 Sampling

Disturbed samples

Disturbed samples are recovered from trial pits and along drilling tools where there is no

attempt to retain the soil constituents. Disturbed samples should however be collected

carefully and placed in airtight tins or jars or in plastic sampling bags. The samples should be

labeled to give the borehole or trial pit identification number, depth of recovery and field

description should be done. The disturbed samples are used for identification tests namely

Field moisture content, PI, grading, compaction and CBR.

Undisturbed sample – cohesive soils

Undisturbed samples are recovered from trial pits and along drilling tools where there is an

attempt to retain the soil constituents. Such a sample is taken in an airtight container with

wax at both ends to prevent moisture from escaping during transportation to the laboratory.

In trial pits the samples can be obtained by pressing a sampler into the ground at the

appropriate depth. The sampler is typically 100mm diameter by 150mm long. In the hand

augur a 38mm sampling tube with a length of 200mm is fitted to the rod after the removal of

the augur. The tube is pressed into the soil and given half a turn to break the soil. The

sampler is then removed and the ends are waxed. In boring rigs a 105mm diameter sampler is

introduced to the borehole to recover a 100mm diameter sample. The sample is usually

381mm long and is fitted with a cutting shoe of about 110mm diameter. The sample is driven

by a falling weight. Any entrapped air or water is expelled from the top through a non return

valve. For soft clays thin walled samplers are preferred to minimize disturbance.

Inevitably there will be some disturbance in the process of retrieving soil samples

from the ground. The least disturbance is for shoes samples cut from the floor of trial pits.

Sample tubes, inserted by pressing, jacking or steady hammering produce some form of

disturbance depending on the thickness of the sampler walls. The degree of disturbance is

related to the area ratio of the sampler tube as given by Equation ****** In general good

samplers have and area ratio not exceeding 25%. Area ratios less than 10% are very good

and are used for very sensitive soils.

x100%



Figure 4.7 Typical sample tubes

Undisturbed sample – cohesionless soils

Various methods have been employed to obtain undisturbed sand samples. These include

freezing, chemical application, and use of compressed air (Smith and Smith, 1998).

Whatever method is employed eventual disturbance occurs as the soil is transported to the

laboratory for testing. In light of these difficulties it is prudent to assess the engineering

properties of cohesionless soils through field testing such as penetration.

Quality class for soil sampling

Table ** below based on Rowe (1972) shows the quality classes for soil samples obtained

from various site investigation operations.

Table 4.2 Quality class for soil sampling

Quality

class

Method of sampling Use of sample

5 Material brought up by drilling tools an no attempt is

made to retain all the soil constituents

Rough sequence of

strata

4 As for 5 but all soil constituents are retained as far as

possible. Bulk an jar samples. Plastic bag samples

Sequence of strata and

remolded properties

3 Pressed or driven thin or thick walled samplers with

water balance in very permeable soils

As above and

examination of soil

fabric

2 As for class 3 above but with water balance all the

time

As 3 and γ, n, mv, cu,

c’ θ’

1 Thin walled piston samplers with water balance As 2 and cv and k

De

Di

Di

De

Sampler tubes Sampler tubes fitted

with a cutting shoe



Borehole logs

Borehole logs summarizes all the laboratory an field tests carried out on samples representing

the various strata encountered in the boring operations. All ground conditions encountered at

the site are also included. The log enables a rapid accurate assessment of the soil profile on a

vertical scale. The details of the various strata encountered including all their geological

formation details which can be inferred are given. The details captured should include the

depth to which ground water was encountered. The description is based on particle

distribution and plasticity based visual inspection and feel. Soil color should also be

recorded.



Courtesy of Norken Engineering Consultants

Figure 4.8 Borehole logs



4.2.4 Scope of Site Investigation

Spacing of the trial pits and or boreholes

The scope of site investigation is dependent on the effect of the construction on the ground.

The scope should be commensurate with the needed geotechnical parameters. Table 4.3

shows the suggested minimum number of borings for the various structures.

Table 4.3 Recommended spacing of investigation trial pits and boreholes

Project Type of soil/Distance between borings Minimum no

Uniform Average Erratic

Multistory 45 30 15 4

1 to 2 storeys 60 30 15 3

Bridge piers and

abutments

30 30 15 1 – 2 per unit

For highways and runways during preliminary design the subgrade soils along the proposed

alignment should be sampled at 1000metres and the samples should be tested to establish the

in-situ CBR, grading and plasticity of the materials. At this stage the material site should be

investigated at 60 meter intervals. In the detailed stage the subgrade is sampled at 500meters

while the material sites are sampled at 30metres.

Depth of investigation

The depth should be such as to capture the geotechnical information needed for the design of

the facility. Equally important is to capture the information needed in the quantification of

the bill of quantities to ensure an accurate specification of the works is carried out. The

recommended depths below the formation of investigation for the various civil engineering

schemes is shown on Table 4.4 based on Figure 4.9 below.

Table 4.4 Depth of investigation

Project Depth In rock Parameters to be established

Column foundations 1.5B-3B 1.5-3m C, θ, N, RQD,TCR

Raft foundations 1.5B 1.5-3m C, θ, N. RQD,TCR

Bridge piers and

abutments

1.5B-3B 1.5-3m C, θ, N, RQD,TCR

Earthworks in fill for

highways

0.5L 0.50m PI, CBR for fill material

Strength of support

Earthworks in cut

highways

0.5H 0.50m Establish the type of excavated

material and strength of support

Pipe works D 0 Investigate type of excavated

material and strength of support



a) Structural foundations

b) Highway earthworks

c) Pipe works

Figure 4.9 Scope of foundation investigations

4.2.5 Site Investigation Reports

List of suitable headings

Title page

Gives the title of the project at a glance

Abstract

The abstract should be approximately 200 words. It is a very important element of the

project and should be prepared with care. It must convey the essence of the site investigation

and all the important findings without ambiguity.

List of contents

Guides the reader to the various chapters

Field work

A brief and complete description of what was done in the field. Boreholes, and trial pits

performed, field testing etc. Actual procedures of standard tests need not be repeated. A

L

L

H

In cut H

D

L

In fill

Retaining walls

B

Piled foundations

B

B

Raft foundations Column

foundations

B



mention of the tests performed is sufficient. New procedures and peculiar fieldwork should

be explained.

Laboratory work

A brief and complete description of what was done in the laboratory work carried out . as in

the case of field testing actual procedures of standard tests need not be repeated. A mention

of the tests performed is sufficient. New procedures and peculiar laboratory equipment and

procedures should however be explained

Site description and geology

An engineering summary of the nature of the site an its geology, including aspects such

excavated areas and what was found, stability of natural slopes, drainage etc

Engineering properties of soils an rocks

A summary of the results of field and laboratory tests and other observations made at the site

Discussion

A reasoned discussion of what design and construction problems are likely to be encountered

in relation to the site and its geological situations.

Recommendations and conclusions

A brief but clear statement of the recommended geotechnical parameters investigated. The

treatment of the various aspects of design should come out clearly and without doubt. Values

of use in design and construction should be summarized viz, allowable bearing capacity,

estimated settlement, suitable types of foundations, construction requirements namely

grouting, compaction etc

References

A list of the books, papers, referred to in the work

Appendices

Appendix A – should contain site plan, borehole logs, photographs, etc

Appendix B – should contain tables of results of field and laboratory test those not included

in Appendix A

Appendix C – Any special or unusual test procedures adopted in the investigation

References:

Craig FR, 1987, Soil mechanics, Van Nostrand Reinhold (International) London

Bowles JE , 1982, Foundation Engineering, McGraw-Hill international book company,

Tokyo.

Tomlinson MJ and Boorman R (1986), Foundation and construction, Longman scientific and

technical, England

Franklin JA and Dussealt MB (1989) Rock Engineering, McGraw-Hill international editions,

London

Chen FH (1975) Foundations on expansive soils, Elsevier scientific Publishing Company

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