28 Agbabian, Higazy, Abdel-Ghaffar

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EARTHQUAKE ENGINEERING AND STRUCTURAL DYNAMICS, VOL. 23,859-876 (1994)

EXPERIMENTAL OBSERVATIONS O N TH E SEISMIC SHEARSUBJECTED TO VARYING AXIAL COLUMN FORCE

PERFORMANCE OF RC BEAM-TO-COLUMN CONNECTIONS

M. S. AGBABIAN*, E. M. HIGAZY' A N D A. M. ABDEL-GHAFFARfCivil Engineering D epartm ent. U niversity of Southern California. Los Angeles, C A 90089-2531, U.S .A .

A N DA. S. ELNASHAI'.'

Department of Civil Engineering, Imperial College of Science, Technology and Medicin e, Imperial College Road, London SW 7 2BU, U.K.

SUMMARYThe paper presents results from the first series of an ongoing experimental study aimed at quantifying the effect of axialcolumn load o n the shear capacity of beam-to-column connections. This is deemed important due to the recent evidenceshowing that vertical earthquake ground motion, when combined with high overturning moments, may cause reducedcolumn compression or even tension. In which case, the concrete contribution to shear resistance in the panel zone isdiminished, which m ay lead to failure prior to the attainment of the full resisting capacity of the beam section. The resultsfirst show that the failure mode of the models was, as intended, shear failure of the panel zone. It is further observed thatthe axial column load has a marked effect on the shear deformation capacity, yield point, cracking pattern, ultimatecapacity and ductility of the panel zone. Differences in the range of 30 per cent in capacity and 50 per cent indeformability were recorded. The preliminary results are useful in providing design guidance for structures ocated inareas of potential high vertical ground motion component. Also, for high-rise structures, where there are large over-turning moments, the results may be of use n ensuring a uniform safety factor (or overstrength) n various non-dissipativeparts of the structure.

INTRODUCTIONModem seismic design utilizes the concept of failure mode control to maximize the reliability and economyof structures. In structural design for gravity and wind loads, overstrength provides enhanced load-carryingcapacity and hence an increase in the safety of the structure. For the case of earthquake-resistant design,however, non-uniform overstrength distributions may result in a brittle failure mode causing low energyabsorption capacity and premature collapse. Moreover, in designing beam-column subassemblages the easeof construction, and hence economy, is of critical importance when estimating the section dimensions, designand detailing. I t is the balance of the overstrength in the panel zone (PZ)region and minimum overstrengthto satisfy economy that makes earthquake-resistant design of the beam-column joint a challenging problem.In reinforced concrete (RC) design, utilizing the concept of weak-beam-strong-column behaviour isimportant, with few exceptions. In all cases, however, it is assumed that the connection will maintain itsintegrity; hence it will continue to transmit loads between beams and columns resulting in the beam-hingingductile failure mode. It is therefore important that the strength of the connection is estimated conservatively,

*Professor.'Research Associate.'Professor.#Reader."Currently Visiting Profe ssor, Civ il Engineering Dep artm ent , Un ivers ity of Southern California, Los Angeles, California, U.S.A .

CCC 0098-8847/94/080859-180 994 by John Wiley & Sons, Ltd. Received 8 April 1993Revised 21 August 1993


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860 M . S. A G B A B IA N E T A L .taking into account all sources of overstrength in the beams and columns. There is, however, a countcr-argument against using large overstrength values for the connection. Over reinforcement will result incongestion, thus resulting in construction difficulties. This, in turn, will reflect on the economy of thestructure. The preceding discussion highlights the importance of estimating accurately the actual capacity ofbeam- column connections, such that a well-controlled overdesign factor is ensured.

INFLUENCE O F VERTICAL GRO UND EXCITATION ON CONNECTION PERFORMANCEAn important parameter influencing the behaviour of beam-column connections, and reinforced concretestructures in general, is the interaction between the response under transverse and vertical ground motion, asdiscussed below.Traditionally, some of the principal codes fail to stipulate that the vertical component of ground motion betaken account of. It is also recommended that, if required, the vertical spectrum is taken as a proportion ofthe horizontal spectrum. This approach is potentially unconservative, due to the following two main points:

(a ) There is a wealth of earthquake records that exhibit a vertical component with a peak ground(b) The frequency content of the vertical motion is often observed to be significantly higher than that of theacceleration well in excess of the corresponding horizontal value, as shown in Table I .horizontal component, as shown in Figure 1.Field observations' as well as results of dynamic analysis of high-rise s t r ~ c t u r e s ~ ~ ~ndicate that thecombined effect of high over-turning moment and excitation of the vertical vibration modes may result in avery significant reduction in the compressive forces obtained from static structural analysis. Some of theresults of a non-linear dynamic analysis conducted by Ko ~ k l e r i , ~re shown in Figure 2. In the cited study, a3-bay and %storey reinforced concrete frame was analysed under different earthquake excitations. Theselected earthquake records had a ratio [between peak vertical and peak horizontal accelerations ( ~ J u , , ) ] f

1.7 or more. Time histories of column axial force presented in Koukleri's study showed very low column axialcompression or even tensile loads for the cases where a combined vertical and horizontal excitation wasconsidered, as shown in Figure 2. Under such conditions, the code-specified concrete contribution to theshear strength is eroded and the total shear force has to be carried by the horizontal reinforcement in thepanel zone. If this frame is designed in accordance with existing code specifications, premature failure of theconnection will ensue, causing sudden loss of transverse stiffness and strength and an increased lateraldisplacement. Hence, design of beam-column connections needs re-examination in the light of the recentevidence that peak acceleration of the vertical ground motion may have a significant effect on the safetymargins of the connections when column loads, varying from high compression to mild tension, areconsidered.

NOTE ON PREVIOUS RESEARCH ON BEAM-TO-COLUMN CONNECTIONSOne of the early studies performed in the field of beam-column connection (Figure 3) testing was by Berteroand P o ~ o v , ~ho tested half scale subassemblages. In their series of tests, inelastic action was to develop in

Table I. Sample earthquakes with high vertical acceleration(Reference 12 )Ear thquake Sta t ion name L7,i1lh

1 El Cen t ro s ta 6 3.862 Westmorland F.S. 2.053 Hollister W areh ouse 2.824 Hollister City Hall 5.445 Salinas Joh n & Wo rk St 2.0


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BEAM-TO-COLUMN CONNECTIONS

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MPERlAL VALLEY EARTHQUAKE. WSTON RD. 15 OCT 1979. 140 DEGPEAK V A L E SA C C X L D U m - -0.398 C ELCUTY - -64.2 cL(/SEC

1 " " I " " I ' ~ ~ ~ l ' ~ ~ ~ l l0.0 2.5 5.0 7.5 r0.0 12.5 15.0 17.5 20.0 225 WO 27 5 30.0 32S 35.0 37 5- _c AI -- W0.0 2 5 5.0 7.5 10.0 125 15.0 V.5 20.0 225 25.0 275 30.0 325 35.0 37 5

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lMPERlAL VALLEY EARTHQUAKE. M T O N RO. 15 OCT 1979. VERTPEAK VALUESACCELERATION - 1665 C VELOCITY - -62.7 CM/SEC

0.0 25 50 7 5 m.0 125 6.0 l75 20.0 225 2 M 216 30.0 323 35.0 375TIME (SEC)

Figure 1. Records of Imperial Valley Earthquake, 1979 (Reference 3)


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Figure 2. Time histories of column axial forces with tension stresses (Reference 3 )


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BEAM-TO-COLUMN CONNECT IONS 863

Figure 3. Typical interior beam-column subassemblage

the beams, as the design was based on strong-column-weak-beam philosophy. They concluded that the earlydrop in resistance and the drastic pinching of the hysteretic loops are caused by bond failure of the beammain reinforcement along the width of the column.

Scriber and Wight,' investigated the aspect of shear strength decay in the connections. The results of theirtest series explained the effect of the percentage of the transverse joint reinforcement on the overall energydissipation capacity, damage control and longitudinal reinforcement buckling. They showed that inter-mediate reinforcement is most effective in improving hysteretic response for doubly reinforced members.

The general shear behaviour of interior, exterior and corner beam-column connections in reinforcedconcrete frames subjected to alternately repeated loading was investigated by Minami and Nishimura.6 Theirtest results showed a tendency of the ultimate shear strength in the panel zone of interior connections to beslightly higher than those of exterior or corner connections.

The influence of reinforcement amount and detailing on the seismic resistance of the connections wasstudied by Goto et al.' as well as Tsonos et a/.' The influence of transverse reinforcement in beam ends andjoints on the connection seismic behaviour was studied by the former group of researchers,' whilst the latter'studied the seismic resistance of diagonally reinforced beam-column connections.

The above was not intended to provide a comprehensive review of previous work on beam-colilmnconnections. The cited examples only serve to demonstrate that the majority of the available test data wereon specimens which did not fail in pure panel zone shear, as elaborated below.

RESEARCH OBJECTIVESIn most of the previous experimental studies4-' mixed mode failure was observed; hence an accurateassessment of the capacity of the panel zone was not possible. In the current test series, several measures weretaken to ensure that the failure mode is purely by panel zone shear failure. This is considered important ifrealistic estimates of the force-resisting mechanisms and the capacity are to be obtained. The latter approachis an extension of the effort of relating the seismic characteristics of the connections to the panel zone (PZ)shear capacity rather than a mixed flexural-shear performance. The low values of column compressive loadused throughout this series of testing simulates a consequence of the vertical ground excitation effect, aspreviously discussed. Hence, under this newly introduced condition, the adopted test philosophy offers anopportunity to reinvestigate the code procedure for the design of RC beam-column connections.


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864 M. S. AGBABIAN E T A L .TESTING PROGRAMME

Specimen detailsAll specimens tested within this programme are one-third-scale models of a multistorey frame jointprototype. The choice of the one-third scale was dictated by the testing machine capacity. A total of three

models were tested as the first set in a series of an ongoing experimental programme. It was decided to haveidentical specimens and to impose strict quality control in order to isolate irrevocably the effect of axialcolumn load. The present specimens were designated S A I , SA2, and SA3. Details of test models are given inFigure 4. The weak-beam-strong-column philosophy was not adopted herein, as it was intended to imposecracking and failure within the panel zone rather than in other elements of the subassemblage.Material properties

The concrete mix was designed according to UBC 2605(d) 3B. Ordinary portland cement type I1 was usedalong with a maximum aggregate size of 0375 in (9.50mm) and a water/cement ratio of 0.76 by weight. Themix had a 5 in (127 mm) slump and yielded a 4 ksi (28 MPa) average strength at 28 days. A 3.00 floz (0355 1)pozzolan 322N was admixed to every 100 lb (45.5 kg) cement. Grade 60 reinforcing steel of minimum yieldstress of 60 ksi (410 MPa) and tensile strength of 75 ksi (520 MPa) was used in all the models. Number 3(10 mm) deformed rebars were mainly used for longitudinal reinforcement of beams and columns. The choiceof the deformed bars was to keep the bar bond characteristics as the simulated prototype. Beam stirrups andcolumn hoops were of number 2 (6 mm) smooth bars.Experimental set-up and instrumentation

A general view of the testing facility is shown in Figure 5(a), whilst Figure 5(b) shows the details of the test-rig. The specimens were tested in a horizontal plane with the boundary conditions designed to provide thenecessary points of contraflexure at the ends of beams and column. The beams and column ends were boltedto steel angles and capped with plates 1/2 in (12.7 mm) thick. A set of links and base plates were used toprovide the necessary boundary conditions. Displacements were imposed on a beam-end mounted on theshake table. The USC-Dual Seismic Shake Table System used herein is a two-station precision, servo-hydraulic testing system designed to operate under software control or direct operator input. Theservohydraulic devices comprise two shake tables equipped with hydrostatic bearings, two series hydraulicactuators and their associated servovalves, two series hydraulic service manifolds and a series hydraulicpower supply. The actuator is of 5.6 kips (f 5 kN) static capacity and a stroke f in (f 0.8 mm). Thetable active controller provides two optional schemes of displacement generation, namely, internal andexternal schemes.The built-in actuator transducer measured the imposed lateral displacement whilst the mounted load cellmeasured the driving force. Four other linear voltage differential transducers (LVDT) were mounted on thebeams and columns at the position nearest possible to the panel zone. The latter were intended formonitoring the joint shear deformation.Strain gauges were placed on the main steel rebars and on selected beam and column stirrups. All rebarand stirrup surfaces were ground to the required smoothness at the appropriate gauge positions. Waterproo-fing with a silicon layer followed strain gauging. Figure 6 shows instrumentation of the test models.Loading regime

In each test, a load-controlled column axial straining was imposed and co-existed with a displacement-controlled cyclic transverse loading on the beams. Whilst the axial column load was kept constant, thetransverse cyclic loading increased every three cycles. To achieve some consistency with previous tests, theECCS (European Convention for Constructional Steelwork) recommended cyclic-loading regime wasimposed. This loading procedure has been extensively used in Europe for testing of beam-columns9 andbeam-column connections." The axial column load levels used were 0 ,5 and 10per cent of the squash load.Cyclic displacement represented the recommended multiples of yield displacement. Plots of the cyclic-displacement regime adopted herein are shown in Figure 7.


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M . S . ACBABIAN ET A L .Control & Computer

& To HydraulicDistribution S ystem Horizontal ActuatorFigure 5(a). General view of the shake-table facility

TEST RESULTSSpecimen S A I

Plot of the applied load versus displacement of the load point for SA 1 is shown in Figure 8. The plot showsan almost linear behaviour during the first three to four cycles up to 0-30 in (7.6mm) displacement. At adisplacement of about 0.32 in (8.13 mm), vertical and horizontal cracks developed through the panel zone(PZ)as shown in Figure 9. These cracks developed at PZ-beam or PZ-column interface and commenced froma corner. Interfaced cracks widened and lengthened as the test progressed, whilst at 0.62 in ( 1 5. 7 mm)displacement, diagonal cracks started developing. The peak load of 2260 lb (1005 kN) was achieved at about0.9 in (22.9mm) displacement when diagonal cracks were completely developed through PZ. Pinching wasclearly observed after the main cracks have developed.In subsequent cycles, there was a gradual and continuing drop of load-carrying capacity. Cracking wasonly limited to the panel zone throughout all stages of loading. Stages of crack propagation andcorresponding displacements are shown in Figure 9. With further widening of the existing cracks, dramaticloss of stiffness and crushing of the joint core, the subassemblage failed at a displacement amplitude of about1.5 in (38.1 mm) and a corresponding stiffness less than one-fifth of the initial stiffness. It is worth noting thatthe specimen stiffness, in all tests, is defined by the slope of the tangents to each cycle of theload-displacement curve. Three further cycles at 1.6 (40.6),1.7 (43.2)and 1.8 in (45.7mm) were applied for theassessment of large displacement residual strength.PZ shear deformation plotted against the applied load is given in Figure 10. Finally, a plot showing therelationship between the percentage of yield strength versus the calculated ductility ratios, as discussed later,is shown in Figure 11 .Specimen S A 2

Specimen SA2 experienced a non-symmetric behaviour during the early stages of testing as shown in theload-displacement plot of Figure 12.A linear response was observed up to a displacement of 0.26 in (6.6mm)where a drift towards non-linearity ensued. A PZ crack that initiated and propagated through th e columndepth resulted in the observed asymmetry. Stages of crack initiation and propagation are depicted in Figure


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868 M . S. AGBABIAN ET A l .

Figure 6 . Ins t rumentat ion set-up, (SA1. SA 2 an d SA3)

( 1 in = 0.0254 m)pPsBVJE

ACQUIRED DATA POINTSFigure 7. Displacement t ime historj

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13 for SA2. The absence of column compressive load led to a rapid crack opening and propagation. PZ crackwidening was then experienced rather than opening of new diagonal or side cracks. The same behaviourcontinued up to a displacement of about 0.65 in (16.5 mm) where another horizontal crack developed acroszthe PZ and through the column depth, as shown in Figure 13. Due to this new crack, symmetrical behaviourwas partially reinstated. Beam cracks were observed firstly at a displacement of 0.30 in (7.6 mm). More crackspropagated through the beam depth and width between displacements of 0.5 (12.7) and 1.2 in (30.5 mm)where stability of crack pattern was reached. A t a displacement of 1.25 in (31.8mm), the peak load of 2150 Ib(9.56 kN) was achieved; PZ diagonal cracks and column horizontal cracks were totally developed. Gradualloss of strength was experienced through subsequent cycles. Cracking spread into the different elements of thesubassemblage rather than being confined to the PZ. Nontheless, the crack concentration was more throughthe panel zone, where crushing leading to failure occurred later. SA 2 failed at a displacement of about 1.97 in(50mm) and a corresponding stiffness of 16 per cent of the initial stiffness. Pinching is clearly manifested inthe load-displacement curve of Figure 12. Joint shear deformation an d ductility ratio plots ar e given inFigures 10 and 14, respectively.


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869

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Figure 8. Load-displacement relation ship of specimen S A l ( 1 in = 0.0254 rn; 1 Ib = 4-448 N)

D , : Ultimate DisplacementSA I - 25% D ,

1

a} Crack Initiation (0.32).A1-75% D ,* SAI- 50% D,*) Crack Propagation (0.85)SAI -100% D ,+} Crack Propagation (1.63 d} Final Crack Pattern (A t Failure)Figure 9. Succe ssive crack patterns, specimen S A l


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0 02 o a-14 0.06 0.08 0.1 0.12Joint Shear Deformation"radians"Figure 10. PZ shear deformation vs. column shear, SA1, SA2 and SA 3 (1 Ib = 4,448 )

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20"w \ II I I I _ _2 3 4 5 6ooy- Ductility Ratio

Figure 1 1 . Ductility ratios vs. percentage of yield strength, specimen SA l

Specimen S A 3Plot of the applied load versus the displacement of the load point of SA3 s shown in Figure 15.The plots

indicated that through the first four cycles and up to 0.42 in (10.7mm) displacement, the behaviour is veryclose to linearity. Departure from linearity started at 045 in (1 1.43mm) displacement. Due to the relativelyhigh column compression in specimen SA3, rack initiation and propagation did not follow the same stagesof its counterparts. As shown in Figure 16 cracking started through the beam and nearest to the point ofdisplacement application. Then beam cracks a t the beam-PZ interface followed at a displacement of 0.39(9.9)o 0.45 in (1 1.43mm). Horizontal PZ cracks through the column depth developed at 0.76 in ( I 9-3mm)displacement, though stayed closed and non-traceable up to 1.0 in (25.4mm) where they started wideningand linking with beam and PZ corner cracks. By further application of displacement, existing cracksdeveloped completely and brought about a stiffness drop to about 35 per cent of the initial value, where thestiffness measured in this study is, as previously defined, the tangent slope to each cycle of theload-displacement curve. Prior to failure, main diagonal cracks opened through the PZ and widened tillfailure. Crushing occurred a t 1.95 in (50 mm) displacement. The specimen ultimate strength (peak load) w as2750 lb (12.23 N) and corresponded to about 1.0 in (25.4mm) displacement. Figure 17 shows the displace-ment ductility ratios plotted against the percentage of yield strength.

The three specimens presented herein, shared the same failure mode which is a pure shear failure inside thePZ with no plastic hinging in the beams and columns framing into the connection. This is attributed to thedesign adopted herein which maintained a flexural strength ratio of 0 8 7 as compared to 1.4recommended bythe codes.


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BEAM-TO-COLUMN CONNECTIONS

I i i , I i-25002 -1.5 -1 -0.5 0 0.5 1 1.5

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SA2- 0% D ,*) Crack Pmpagalion (0.85)

c) Crack Propagalion (1.45) d) Final Crack Pattern (A t Failure)Figure 13. Successive crack patterns, specimen SA2


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872

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M . S. AGBABIAN ET A L .

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I I I I 12.5 3 3.5 4 4.5 5Ductility RatioFigure 14 Ductil i ty ratios vs percentage of yield strength, specimen SA2

I I I I I I 12 2.5 3 3.5 4 4.5 5100; 1.5 Ductility Ratio

Figure 14. Ductil i ty ratios vs . percentage of yield strength, specimen SA2

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Lateral Displacement "in"Figure 15. Load-displacement relationship of specimen SA 3 (1 in = 0.0254m; 1 Ib = 4.448 N )

ANALYSIS AND DISCUSSION OF TEST RESULTSStrcngth analys i s

As previously mentioned, the subassemblages were designed according to the ACT 3 18-89 code provisionswith several measures to ensure a pure PZ failure. Hence, a more precise estimate of the PZ shear capacity is


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BEAM-TO-COLUMN CONNECTIONS 873

SA3-50% D ,I 1 -I t

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a ) Crack Iniliation (0.42) b) Crack Propagation (0.84)

SA3-75%Du SA3-10O%Du

Uc) Crack Propagalion (1.5) c ) Final Crack Pattern (At Failure)

Figure 16. Successive crack patterns, specimen SA3

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. . . _ . . _ . . . _ . . _------j..........................-i I I I I100,v 2 3 4 5 6Ductility Ratio

Figure 17. Ductility ratios vs. percentage of yield strength, specimen SA 3

obtained. It should be noted that the experimentally acquired capacity represents a pure shear failure case,where all the possibilities of a mixed failure mechanism were eliminated. The acquired capacities wereobtained from the recorded column shear along with the corresponding measured strains in the beamlongitudinal reinforcement. The influence of axial column compression was controlled through the imposedstrain level in the column reinforcement.On the other hand, the analytical capacities were obtained from a proposed simple mechanical model. Thisis based on the two basic mechanisms contributing to the shear transfer through the panel zone, namely, thediagonal strut mechanism and the joint truss mechanism. The model formulation considers the ultimatecapacity of the PZ transverse reinforcement to be the contribution of the joint steel to the overall shearcapacity, whilst the concrete contribution is represented by its resistance to the panel zone diagonalcompression. Finally, the influence of the axial column compression is accounted for by evaluating thedimensions of the resisting concrete strut as a function of the level of the column normal stress. The analyticalmodel formulation and verification are given elsewherel .Values of the acquired and analytical joint shearcapacities along with the column axial compression are given in Table 11.The effect of the column compression on the joint shear capacity is manifested in the difference of more than19 per cent in the capacities. The corresponding analytical difference is 14 per cent for the same variation


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874 M. S. AGBABIAN ET A L .Table 11. Recorded and calculated shear capacity

Axial colum n Joint shear capacitycompression(YoPu) Experimental-kips (kN ) Analytical-kips (k N )0 22.07 (98.17) 20.74 (92.25)5 24.16 (107.45) 21.53 (95-77)10 27.23 ( 1 2 1 . 11 ) 23.92 (106.40)

in the axial load. A difference of about 6-10 per cent (well within experimental e rro r margins) between theexperimentally acquired strength an d the analytical strength is observed which show s that the model used iscapable of predicting reasonably representative strength estimates.Shear deformation analysis

A key parameter in the analysis of PZ performance is the joi nt (PZ) shear deformation. Figure 10 shows therelationship between the column shear an d the corresponding shear deform ation at different levels of columncompression. It is shown that a slight increase in column compression results in a relatively high increase inthe resistance of the panel zone to shear deform ation. Hence, a lower interstorey drift would be experiencedand there would be higher chances of maintaining the structural integrity. This will also have a beneficialeffect on non-struc tural components. Conversely, the absence of column compression increases substantiallyth e PZ shear deformation with a factor of 50 per cent or more, thus increasing overall displacements of thestructure.Moreover, Figure 10 shows tha t specimens with 5 and 10 per cent axial column compression followed thesame pattern of progress with respect to shear deformation up to the peak load of the former. Duringsubsequent loadin g stages, and du e to the role of the relatively high com pression in crack closure, the lattershowed a better shear deform ation resistance. Such behaviour is attributed to differences in the extent of PZcracking in the models during post peak load cycles. The same argu men t holds for the shear deformation ofspecimens with no axial column compression and that with 5 per cent compressive load. Considerabledifferences in the behaviour are observed from Fi gur e 10up t o the level of com plete development of cracks inboth specimens. Afterwards, the role of axial load in crack control diminished and the two specimensdeformed identically u p t o failure. Table I11 gives the values of the PZ shear deformation corresponding tofirst yield and failure, respectively, along with the PZ shear deformation ductility Rpz.It is observed from Tab le I11 that a 53 per cent decrease in the local shear d eform ation ductility is caused bya 10 per cent axial column load reduction. This is attributable to the significant effect of the columncompression on the yield shear deformation.Ductility analysis

In the present analysis, two different overall ductility ratios are presented. These a re based on (i) the localshear deformation ( R , )and (ii) the global displacement ( R 2 ) . t is worth noting th at the form er ductility ratio

Table 111. PZ shear deformationsAxial columncompression PZ shear deformation (radians)

(% Pu ) At yield At failure R ,,0 0.056 0-121 2.165 0-039 0.119 3.0510 0 0 2 5 0.1 16 4.64


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BEAM-TO-COLUMN CONNECTIONS 875Table IV . Ductility and energy absorption capacity

Axial column Energy absorption Ductility ratioscompression capacity(YoPu) Ibin (Nm) R , R2~~

0 2775.0(313*5) 3.95 4 9 85 4345.0 (490.89) 4.10 5.1910 4537.5 (512.64) 5.40 5.75

is based on the cumulative shear deformation plot while the latter is based on the overall load-displacementcurve. For a better understanding of strength degradation, the percentage of yield strength versus the localand global displacement ductility is depicted in Figures 1 1 , 14 and 17.The yield point was estimated from a proposed mechanical model based on the joint shear resistancemechanisms, namely, the truss and the diagonal strut mechanism along with the internal force diagrams ofthe joints." In Reference 8, it was observed that a strength reduction is recorded corresponding to themaximum ductility attained. This is at variance with the current results, where no such clear reduction isobserved. This emphasizes the exactitude of the current approach, where the failure mode is that of shear inthe PZ. A clear drop in load-carrying capacity would only be recorded if and when the horizontal steelexceeds its ultimate stress. In the current test series, the steel strain was well below the strain at ultimatestress.Being more contentious to estimate, the failure point was determined using two different approaches. Thefirst is the displacement corresponding to a fixed percentage reduction of the ultimate strength whilst thesecond is the failure point corresponding to a given local ductility. It should be noted that in all cases, thelocal criterion is mapped onto the overall load versus displacement curve of the subassemblage in order toevaluate the overall ductility. Ductility ratios R , and R , for the tested specimens along with the energyabsorption capacities are given in Table IV.It is clearly demonstrated in Table I V that a decrease of 10per cent in the column axial compression resultsin a loss of displacement ductility in the range 14-26 per cent based on the first and second approaches toductility, respectively.

CONCLUSIONSCurrent seismic design practice recommends that the vertical components of earthquake ground motion maybe represented by the horizontal spectral values scaled by about 70 per cent. This ignores the existingdatabank of earthquake records where vertical peak ground accelerations may be well in excess of thecorresponding horizontal values." It also ignores the fact that the frequency content of the verticalcomponent is invariably different from that of the horizontal component.In recognition of the above, recent analytical studies have focused attention on the possible consequencesof high over-turning moments and co-existing vertical excitation. It was shown that reduced compression, oreven tension, may be experienced by intermediate stories of medium to high-rise RC structures. Consequen-tly, the shear resistance mechanisms of beam-column connections, usually considered as concrete and steelcontributions, should be re-examined. This is the motivation behind this work.Results from the first three tests of an ongoing research programme at the University of SouthernCalifornia are reported in this paper. The models were designed to exhibit a failure mode entirely controlledby the panel zone. The models were tested under variable amplitude cyclic displacement-controlled loading.Axial load levels of 10, 5 and 0 per cent of the squash capacity were applied to thc identically reinforcedspecimens which were manufactured under very strict quality control.The results presented in the paper indicate that the behaviour of the panel zone is clearly and significantlyaffected by the axial column load. The overall displacement response of the sub-assemblage decreased by 22per cent for a decrease in the axial load from 10 to 5 per cent of the squash load. Moreover, the displacement


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876 M. . AGBABIAN ET A L .Table V. Summary of test results

Axial column PZ shear Ductility ratioscompression capacity(/.I kips (kN) R , R , R , ,0 22.07 (98.17) 3.95 4,975 2 1 65 24.16(107 .46) 4.1 5.19 3.0510 27.23(121.11) 5-4 5.75 4.64

ductility ratios (calculated by two approaches) have shown a decrease of 21.5 per cent for the same range ofvariation in axial column load. The local shear distortion ductility ratio also decreases by a wide margincorresponding to a decrease in axial load. Finally, the overall capacity of the subassemblagedecreased by 19per cent from the 0 per cent axial load specimen to the 10per cent case. The results obtained are summarizedin Table V.

The implications of the above results are that the concrete contribution to the shear resistance of panelzones of RC beam-to-column connections is sensitive to the level of axial column load, which is in turn afunction of the overturning moment and the vertical component of earthquake ground motion. In areas ofpotentially high vertical component, the concrete contributim cannot be relied upon, and modified designprocedures are warranted. For reasons of economy, and to avoid steel congestion in the panel zone, well-controlled overstrength margins are required. These can only be evaluated if a pure panel zone failure modeis imposed, as in this test series.

REFERENCES1. N . N. Ambraseys, Private communication, 1990.2. 0. Papadopoulou, Effect of vertical ground motion on the response of multistorey reinforced concrete structures, M.Sc.Dissertation, Imperial College, University of London, September 1988.3. S. N. Koukleri, The effect of vertical ground excitation on the response of R.C. structures, M.Sc. Dissertation, Imperial College,University of London, August 1992.4. V. V.Bertero and E. P. Popov. Hysteretic behavior of ductile moment-resisting reinforced concrete frame components, Report No.EERC 75-16, University of California at Berkeley, California, April 1975.5. C. F. Scriber and J. K. Wight, Delaying shear strength decay in R/C flexural members under large load reversals,Proc. 7th WCEE.Vol . 7 , Istanbul, 1980, pp. 31-39.6 . K. Minami and Y.Nishimura, Hysteretic characteristics of beam to column connections in steel reinforced concrete structures,Proc. 7th WCEE, Vol. 7, Istanbul, 1980, pp. 305-309.7. Y.Goto, 0. Joh and T. Shibata, Influence of transverse reinforcement in beam ends and joints on the behavior of R/Cbeam-column subassemblages, Proc. 9th WCEE, Vol. 4, Tokyo/Kyoto, 1988, pp. 585-591.8. A. Tsonos, I. Tegos and G. Penelis, Seismic resistance of type 2 exterior beam-column joints reinforced with inclined bars, A C Istrucf . . 89. 3-12 (1992).9. A. Y.Elghazouli, A. S.Elnashai and P. J. Dowling, Experimental behavior of ductile partially encased composite beam-columns,Proc. earthquake, blast and impact: measurement and effects o vibration conj, Manchester, Society for Earthquake and CivilEngineering Dynamics, September 1991, pp. 21 1-220.10. G.Ballio and F. Perotti, Cyclic behavior of axially-loaded member; numerical simulator and experimental verification,J. construcr.steel res. 7 ,3-41 (1987).

1 1 . E. M. Higazy, Seismic shear performance of beam-column subassemblages in multistory RC structures, Ph.D . Dissertation, CivilEngineering Department, University of Southern California, Lo s Angeles, California, USA., April 1993.12. N. A. Abrahamson and J. J. Lithehiser, Attenuation of vertical peak acceleration, Bull. seism. soc. Am. 79 , 549-580 1989).

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